Behaviour of extended end-plate connections under cyclic loading

Behaviour of extended end-plate connections under cyclic loading A. Ghobarah, A. Osman and R. M. Korol Department of Civil Engineering and Engineering Mechanics, McMaster University. Hamilton. Ontario, Canada LSS 4L7

The results of five tests on bolted end-plate beam-to-column connections are described. The specimens were subjected to cyclic loading simulating earthquake effects on a steel moment-resisting frame. The objective of the work is to determine the behaviour of this type of connection under cyclic loading well into the inelastic range and to ascertain the effect of design parameters such as end-plate thickness, column flange stiffener and bolt pre-tension force on the overall behaviour. Observations are made concerning the response of the connection and its elements. Information and implications on the design of this type of connections to sustain simulated severe earthquake ground motion are presented. It is concluded that properly designed and detailed extended end-plate connection can provide excellent ductility as moment-resisting components in the seismic design of frames. K e y w o r d s : earthquake, steel, connection, frame, end-plate, test In steel moment-resisting framcs subjected to earthquake type loading, the beam-to-column connection constitutes a particularly critical region, since large amounts of energy must be dissipated with a minimal loss of strength or stiffness. In the last two decades, study of the behaviour of different types of beam-to-column connection subjected to seismic loading has received considerable attention. Most of the work focused on studying the behaviour of fully welded or partially bolted connections =- ~ Recently, the steel fabricating industry has opted towards fully bolted connections for field work. A connection that is gaining popularity is the extended endplate connection. In this type of connection, an end-plate is shop-welded to the beam which is then field-bolted to the column flange, as shown in Figure 1. This connection type offers several advantages such as low cost, simple and convenient erection procedures, and good quality control since field-welding in adverse environments is undesirable in the best of conditions. However, from an analysis and design point of view, sufficient information about the distribution of stresses and forces in end-plate connections is frequently lacking. The state of stress in such a connection depends not only on the bolt's arrangement, the weld type and its distribution, but also on the relative deformation of the connecting elements. Tests conducted on such end-plate configurations under monotonic loading have revealed that this type of connection can indeed provide reasonable strength, stiffness and adequate ductility 4'5. However, there seems to Ix a dearth of knowledge about the response of such connections to cyclic loading. Some

initial work has been done in this regard. Pilot tests conducted by Walpole in 19816 and Ballio et al. in 19857 on this type of joint with reversed loading do show promising results. Walpole tested four connections in which two different types of column flange stiffening systems were employed. The first is the common extended beam flange stiffener type that consists of plates filletwelded to both the column flanges and the column web. The second type has the plate stiffeners positioned parallel

Figure 1 Typical extended end-plate connection

0141-0296/90/0100 ! 5- ! 3/$03.00 1990 Butte=worth & Co (Publishers) Ltd

Eng. Struct. 1990, Vol. 12, January

15

Extended end-plate connections under cyclic loading: A. Ghobarah et al. to the column web and welded to the column flange tips. Only minimal experimental data regarding the behaviour of these connections were obtained from these tests. In the research by Ballio et al., four connections were also tested. In three of them, beam flanges and webs at the locations of the expected plastic hinges were stiffened to prevent local buckling of the beam. Thick end-plates and column flange stiffeners were employed for the four specimens. The main objective of this work was to develop a model to simulate the overall behaviour of the connection under severe earthquake. It is evident from these earlier investigations that little attention was paid to the behaviour of the individual components of the connection, and, thence, their contribution to the overall joint behaviour. Therefore, the objective of this study is to investigate the response of extended end-plate connections in detail under conditions of severe cyclic loading. In particular, it is planned to assess the following: (!) the behaviour of unstiffened end-plate connections (this type represents a baseline for comparison and is the minimum cost joint); (2) the behaviour of stiffened extended end-plate connections; (3) a modification type that involves stiffening the endplate itself; (4) the individual behaviour of the connecting elements that include the beam, column flanges, end-plate, bolts and stiffeners. To undertake the proposed test programme, a special test frame was designed and constructed. In all, five end-phtte beam - column connections were tested under cyclic loading. Measurements of strains, displacements and rotations of the various components of the connections were recorded during the loading regime. Details of the testing programme and results are now presented.

Experimental programme Conneelion design In the design of moment-resisting frames under severe lateral loads, it is reasonable to assume that the points of inflection are located at the mid-span of the beams and the mid-height of the columns. A simple cantilever type beam-column connection, such as the one shown in Figure 2, was chosen as the subassemblage for this study. The cantilever length represents approximately one-half the length of typical beams in a moment-resisting frame. For simplicity of testing, no attempts were made

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16

Eng. Struct. 1990, Vol. 12, January

Load

to simulate axial force in the column; attention was principally confined to a study of the behaviour of the connection itself. The material used for all five test specimens including beams, columns, continuity flange stiffeners, end-plates and reinforcing stiffeners was in accordance with CSA G40.21-M300W steel (equivalent to 44 ksi structural grade). The high tensile bolts used are specified as 25 mm diameter, grade ASTM A490M. For all welding, compatible electrodes were used (CSA W48.1). For the tested specimens, a W360 x 45 (W14 x 30) section was used for beams and W360 x 79 (WI4 x 53) or W360 x 64 (WI4 x 43) section for the column stubs. The half-flange width-to-flange thickness ratio, b/t, for the beam section was found to be 8.72, with the web depth-to-web thickness ratio, h/w, equal to 42.57. These ratios meet the Canadian code requirements a and L R F D (Load Resistance Factor Design) requirements 9 for beams in highly seismic areas. On the other hand, according to the new provision of the New Zealand standard for the design of steel frames t°, this beam section is only capable of providing limited ductility.

Experimental arrangement In the test set-up, the column stub was rigidly clamped to a rigid fixture frame by sixteen 25 mm diameter grade A490M bolts. The fixture frame was pretressed to a 600 mm thick floor slab of the laboratory by four 50 mm diameter rods. A hydraulic actuator was used to apply the load to the beam tip in the vertical plane. The actuator was fixed to the floor by a pair of stiff angles pretressed to the floor slab by two 38mm rods. The general arrangement of the set-up is shown in Figure 3 with a pictorial view of a test shown in Figure 4. A clevis welded to a 50 mm diameter high-grade steel rod was bolted to the end of the beam by a 50 mm diameter pin; the other end of the rod was connected to the actuator. Two angles were installed near the end of the beam to provide lateral support to prevent lateral buckling. Since in actual practice the top flange of a beam is generally supported laterally by the floor deck, a guide to prevent lateral displacement of the top flange was provided at the mid-span of the cantilever.

Test specimen description In order to maintain consistency, the five specimens were fabricated in the university workshop by the same fabricator. The specimen consists of a beam connected to the column stub by the joint. The column's stub web was reinforced by 8 mm doubler plates. All beams were welded to the end-plates by 10 mm and 7 mm fillet welds for the flange and web, respectively. These sizes are in accordance with CSA-W59 and were computed to provide sufficient fastening strength to develop the fully plastic moment of the beam. All welds were visually inspected and repaired if any defects were discovered. Bolt holes were drilled 2mm larger than the 25mm nominal diameter of the bolt. Direct Tension Indicating washers were used underneath all bolt heads to check the bolt tightening force tt. Coupons from the beam sections were extracted to assess the member's physical properties. Beams A-l, A-3 and A-4 were cut from the same l-beam and beams A-2 and A-5 were taken from

Extended end-plate connections under cyclic loading. A. Ghobarah et aL

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Eng. Struct. 1990, Vol. 12, January 17

Extended end-plate connections under cycfic loading. A. Ghobarah et el. deformations in the end-plate. Again, four continuity plates were provided to stiffen the column flange (Figure 5c').

another I-beam. The test results are listed in Table 1. Details of the test specimens are shown in Figure 5.

Spec'mzen ,4-I. The specimen was constructed by welding Specimen A-4. The thickness of the end-plate was taken

the beam to a 25 mm thick end-plate. The end-plate thickness was designed to sustain 1.3Mp of the beam, where Mr, represents the plastic moment of the beam calculated based on the coupon yield stress, to allow for the strain hardening for this beam section. No continuity plates were used to stiffen the column flange. The column stub was a W360 x 64 section (Figure 5a).

to be 19 mm which was chosen to sustain Mp of the beam, No continuity plates were provided to stiffen the column flange. During the welding process, a small degree of end-plate bending was observed as shown in Figure 6. Tests at both Vanderbih University and at the University of Oklahoma tz in which similar connections were subjected to monotonic loading showed that such end-plate distortion did decrease the overall connection capacity but did subject the weld to high stresses during the field bolt-tightening operation (Figure 5d).

Specimen A-2. The specimen was made by welding the beam to a 25mm thick end-plate. Four 9 m m thick continuity plates were used to stiffen the column flange. The continuity plates were attached using 6 m m allaround welding (F(~ure 5h).

Specimen A-5. The spccimcn was fabricated by welding the beam to a 16 mm thick end-plate, Continuity plates were provided to stiffen the column flange. Also, stiffener plates were welded to the end-plate and beam flanges. To avoid distortion of the relatively thin end-plate during welding, a special fabrication technique was used. The

Speehm'n A-3. The specimen was fabricated by welding the beam to a 19 mm thick end-plate. In this case, two 9 mm thick stiffeners were positioned and welded to both the end-plate and the beam flanges to avoid large Table I

Mechanical properties of tensile coupon specimens

Specimen no.

Coupon location

Yield stress (MPa)

Tensile strength (MPa)

A-1

Flange Web

310.9 315.7

5000 480.7

A-2

Flange Web

316.1 322.1

503.3 480.6

A-3

Same as specimen A-1

A-4

Same as specimen A-1

A-5

Same as specimen A-2

,l'~/

Distorted, end-plate

Beam

%,, Figure 6

Distortion of end-plate due to welding

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Eng. Struct. 1990, Vol. 12, January

Section 5

Extended end-plate connections under cych'c loading. A. Ghobarah et al. end-plate was first welded to the beam using 4 mm fillet welds. The stiffener was then tack-welded to the end-plate and beam flange. Foilov,'ing this stage, thc welding process was completed, it is believed that this technique prevents the end-plate distortion, facilitates the erection and does not affect the connection strength IF(eure be).

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[.otld#lg s('qtl('lt¢'(" To simulate seismic fi)rccs, the test specimens were subjected to a quasi-static cyclic loading. Although the severe lateral loads considered in this study were of a dynamic nature, it is believed that the rate of strain developed in the structural system was not sulliciently high to introduce significant variations in the so-called "static characteristics" of the material ~'t. Before reaching the yield point, an individual specimen was subjected to four load cycles of half the expected yield value, The behaviour of the connection was therefore anticipated to be elastic. This pretest technique was employed to ensure proper connection set-up, welding, bolting and to check the data recording devices fi)r proper functioning. The load was then increased until the initial beam yielding was recorded by the strain gauges. Two cycles of reversing loads of this magnitude wcrc then slowly applied. For subsequent loading cycles, the beamtip displacement was incrementally increased by half the yield displacement up to partial ductility of four (partial ductility is defined as the ratio of beam-tip displacement to the beam-tip displacement at the first yield). If no hulure was detected two additional cycles, one at five and another at six partial ductility, were applied. After that, the test was terminated since at this level a high ductility was achieved, in some cases, the test was terminated before reaching partial ductility of six due to scvcrc lateral

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For monitoring the actuator load, a special built-up load cell was used. The cell consisted of eight strain gauges attached to the actuator rod (four transversely and four longitudinally). The load cell was calibrated after each test in both tension and compression on a t, niaxial Tinius machine. For measuring the beam-tip displacement a cable-type Linear Voltage Displacement Transducer ILVDT) was used. Dial gauges and I.VDTs were used to measure the end-plate rotation and the column flange rotation. Strain gauges wcrc used to monitor the onset of beam flange local buckling and to determine initi:d yielding of the beam. For the first two tests (A-I and A-4), two bolts in each specimen were instrunlented using strain gauges to monitor the fluctt,ation in the bolt pro-tension fl)rcc. In subsequent tests, fl~ur bolts were strain gauged in each specimen. The boll force was measured by the method developed by Suttees and Ibrahim~ L In this method, the average strain in a rosette gattge glued to the bolt head was calibrated against the fi)rcc in the bolt. A special rig wus fabricated to calibrate the bolt force prior to the test on a uniaxial Tinius testing machine. All data from strain gauges and transducers wcrc scanned by a naultichanncl Autodata ~1 Scanner System. The readings were recorded using a micro compt,ler system and were also punched onto a paper tape device to provide backup.

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torsional buckling of the beam. A typical loading routine is presented in Fi.qure 7. A typical loading sequence of a test specimen consisted of 10 inelastic cycles. During the test, the loading was temporarily stopped at each 5 mm displacement increment to allow for scanning the stram gauges and transducers by the data acquisit,on system.

E x p e r i m e n t a l results 3"he data recorded during each test can be grouped into the following categories: (I) beam load versus beanl-tip displacement; (2) beam nlomerlt versus column flange rotation in tile case of the unstiffened A-I and A-4 connections; (3) be;.tnl inonlcnt versus end-plate rotation and bolt ex t e n s i o n

(4) variation of bolt pre-tensioning forces. Strains at various other locations were also recorded. in particular, special attention was focused on the strains in the end-plate stiffeners.

The test data were assembled under the various categories to illustrate certain aspects of the behaviour of a particular specimen. Also, such a categorization allowed determination of the influence of a particular variable on the overall bchaviour of the subassemblage. In designing the experiment, special attention was paid to the measurement of the different types of data. It is of interest that the contribution of each element in the connection to the overall bchaviour could thus be isolated. Since the column stub was rigidly fixed to the fixture frame, the end-beam deflection represents the total displacement duc to elastic and inelastic deformations of the beam, column flanges, end-plate and bolts. In the study, column flange rotations were recorded fi~r the unstiffcncd connections. If these rotations are multiplied by the cantilever length and then a subtraction made from the overall displacement, the resulting displacements arc due to a combination of bolt extension and beam and end-platc deformations. Then, by recording the rotations of the end-plate relative to the column and multiplying them by the cantilever length, the contribution of the end-plate rotation and bolt extension was determined. What remains will be the elastic and inelastic beam rotation, which could then bc separated.

Eng. Struct. 1990, Vol. 12, January

19

Extended end-plate connections under cycfic loading." A. Ghobarah et al. was observed during latter cycles of the test. it was evident that column distress could have been avoided with the use of column stiffeners. It is perhaps significant that the column flanges of this specimen met the design recommendations of Mann and Morris t s. Separation between the column flange and the end-plate was noted up to tile end of the test. Meanwhile, gradual degradation in the bolts' pre-tcnsion forces were recorded. Failure of the specimen was attributed to excessive lateral displacement of the column flange, as is evident from Figure 9. It was contirmed that most of the energy was dissipated in the

Test rcszdtx In the following, the bchaviour of each specimen is examined in detail. The hysteresis loops of the beam-tip deflection versus beam-tip load arc presented. Observations regarding the behaviour of individual elements in the connection and the failure modes are introduced. Specimen ,4-1. The hysteretic loops of the beam-tip load versus beam-tip deflection for specimen A-1 arc shown in Figure ,~', Well behaved hysteretic loops can be observed. However, severe damage to the column flange 20O

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Specimen

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Eng. Struct. 1990, Vol. 12, J a n u a r y

I 150

200

Extended end-plate connections under cyclic loading." A. Ghobarah et al. Specimen A-3. The hystcretic loops of the load versus beam-tip deflection for specimen A-3 are shown in Figure 13. The loops exhibited stable characteristics up to the buckling of the beam flanges and web. The presence of the end-plate stiffener had the effect of shifting the plastic hinge location away from the face of the connection to the section where the triangular stiffeners terminate. The test was terminated after a full ductility of 8.5 was reached. It is evident from Figure 14 that the specimen failed through lateral torsional buckling. No damage to the column flange or the end-plate stiffener was observed and as such most of the energy was dissipated through inelastic action of the beam.

column flange with a minor amount dissipated by the Ix'am flange.

Specimen A-2. The hysteretis loops for this specimen consistently exhibited stable characteristics as shown in Figure !0. However, as the loading progressed, buckling of the flanges and the web of the beam caused deterioration in the load-carrying capacity of the connection. During the test, reduction in the bolt pre-tension force was noted. Tip displacement due to bolt extension and end-plate rotation is shown in Figurt" 11, with the failed specimen presented in FL~ure 12. The test was terminated after a pronounced ductility of9.18 was reached. Analysis shows that most of the energy was dissipated through inelastic action of the beam. Z00

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Eng. Struct. 1990, Vol. 12, January

21

Extended end-plate connections under cyclic loading. A. Ghobarah et aL

Figure 12 SpecimenA-2 after failure

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designed compared with the Packer and Morris recommendations ~". ()il Ihe olher h;ind, ii is overdcsigned based oil Kri~.hll:inlurlh)"s work l" Meanwhile, the coltilnn tlange met the rcqtiircments designated as conservative b) Kulak and Fisher TM. !l is app;.ircnl [ronl the hvsieretic loops in F&,ure 15 Ihal lho response was initially well behaved. During the fourth cycle, however, a clear:age crack was initiated at the toe of the weld between the end-plate and the beam llangc ;.|lid propagated in subsequenl cycles Ihrough the end-plate thickness. "lhe

22

Eng. Struct. 1990, Vol. 12, January

fracture or the extended parts of thc end-plate i:;itisctl a sudden drop in lhc load-carrying capacity of the Ctillncclion. tlowevcr, this failure did not result in the colnplctc collapse of the Collncction, since the inside bolts could slill sustain a significant load. No inelastic aclion was observed in the bc;.inl. Meanwhile, scvcrc dam;igc was observed in the column tlangc as shown in Figurc 16. The analysis shows that most of the energy was dissipated in the end-plate and in the ¢ohimn flange while the beam participated very little in the energy dissipation proccss.

Extended end-plate connections under cycfic loading." A. Ghobarah et al. connection in dissipating energy. During the test, strain records taken over the end-plate stiffener show that the stiffener had yielded. Near the end of the test, crack initiation at the toe of the weld between the beam flange and the end-plate stiffener was observed. Again, a reduction in the bolt pre-tension forces was recorded. The loss of pre-tension force for one of the bolts as measured is shown in Figure 18. Failure of the specimen is shown in Figure 19 with the arrow indicating the crack that developed at the toe of the weld between the end-plate stiffener and the beam flange. Discussion of test results

Figure 14

Specimen A-3 after failure

Specimen A-5. The hysteretic loops of load versus beamtip dellection for specimen A-5 are shown in l"(~ure 17. As in specimen A-3, beam flange and web buckling were responsible for the deterioration of the loops, it was observed that the web buckling occurred after six cycles, cotnpared with specimen A-3 when it occurred after live cycles. This can be attributed to the participation of the

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The beam load capacities and ductility levels as measured during the tests are listed in Table 2. Examination of the hysteresis loops of Figures 9, 12 and 16 shows their remarkable stability, in spite of the decay in the loadcarrying capacity, excellent ductilities were obtained for specimens A-2, A-3 and A-5. The decay in the load is associated with beam flange/web buckling. It is likely that the rate of the decay in strength could be reduced by using more conservative slenderness ratios for both the flanges and the web (i.e., reduced hit and h/w ratios). Connections with unstiffened columns A-1 and A-4 showed very poor behaviour compared with those stiffened. It is suggested that such a connection can be used when limited ductility is adequate. Degradation in bolt pre-tension forces w~,s observed to continue as the magnitt,de of the cyclic load increases. Since the ductility demands of the beam results in bending moments approaching 1.3 M~,, it is recommended that the bolts be designed for a h~rce corresponding to such a moment. This will h;,ve the effect of avoiding the failure of the bolts and the loss of the pre-tension force. The results also show that the end-plate thickness can be reduced significantly if it is properly stifl~:ned. This approach also would facilitate erection, since a thin end-plate can be clamped into contact more easily than

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Eng. Struct. 1990, Vol. 12, January

23

Extended end-plate connections under cyclic loading. A. Ghobarah et aL

NO. A,I

AFTF.I{

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24

Eng. Struct. 1 9 9 0 , Vol. 12, J a n u a r y

50

100

150

Extended end-plate connections under cycfic loading. A. Ghobarah et aL Table 2 Joint capacity and ductility P3oo

P'v

Pv

Pu

(kN)

(kN)

(kN)

(kN)

A- 1

96.9

100

99

147.7

4.65

3.93

A-2

96.9

102

102

154.5

9.18

6,41

A-3

101.7

105

111

156.4

8.52

5.35

A-4

96.9

100

101.1

134.6

5.28*

3.64"

A-5

101.7

107

110.3

1 58.9

9.54

5.71

Soeclmen no.

P30o

F

P

Nominal cantilever tip load at yield stress 300 MPa Yield cantilever tip load based on the coupon test

P,

Measured tip beam load at yield

Pu

Maximum attained load during the test

F

Full ductility (ratio of deformation at the beam tip measured from the position of zero loading to maximum displacement during the last cycle divided by the deformation of the beam tip at the first yield, p~/& v as shown in Figure 17) Partial ductility (ratio of deformation at the beam tip measured from the position of zero displacement to maximum displacement during the last cycle divided by the deformation of the beam tip at the first yield, p2/Av as shown in Figure 17)

*Specimen failed during test (results obtained after failure of the end-plate)

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Figure 18 Variation of bolt no. 1 pre-tension force with loading steps (specimen A-5)

thick end-plates. Indeed, for large beams, unstiffcncd end-plates would necessitate very large thicknesses since they should bc designed to sustain 1.3.tlp of the beam.

Conclusions Bztscd on this study, some preliminary conclusions and design code implications may be suggested. 11) Based on the results of spccimcn A-I and A-4. the use of an tmstiffcned column at the connection is not recommended from the seismic design point of view. 12) Pre-tension forces in all the bolts showed degradation with repeated load cycles. Thc drop in the pro-tension fi)rcc continues with increasing the load. To ensure

that the bolts do not fail and do not lose their pre-tension forces signilicantly even during moderate earthquake excitation, it is suggested that the bolts bc designed to sustain a force corresponding to beam moment of 1.3Mp. (3) In the case of unstiffcncd end-plates, the end-plate must be designed to sustain 1.3Mf, of the beam, this provision allowing for the strain hardening. Using the current code limitations will likely result in overly thin cnd-platcs which arc expected to rail during severe earthquakes. (4) In case of stiffened end-plates, their thicknesses can bc designed to sustain M~, of the beam. It should be noted that reducing the end-plate thickness more than

Eng. Struct. 1990, VoI. 12, January

25

Extended end-plate connections under cyclic loading: A. Ghobarah et aL

Figure 19

Specimen A - 5 at failure

this might result in increasing the load transmitted through the stitTener and might cause fracture of the weld between the stiffener and the beam Ilange. (5) It is suggested that the same fabrication technique used in fitbricating the connection of specimen A - 5 - welding the end-plate to thc beam with 4 mm lillct weld and then tack-wclding the cnd-plate stiffener to the beam and the end-plate bc used with stiffened end-plates. This techniquc will facilitate the erection and reduce the end-plate distortion which will reduce the stresses on the weld during tightening of the bolts. (6) T h e Canadian requirements for beam h/t and h/w ratios fi)r highly seismic areas (CA N 3-S 16. I proposal fi)r clause 26) seem nonconservative and may lead to reduction in the h)ad-carrying capacity of the beam. (7) In all the tests, the connections were able to sustain moments higher than the beam's nominal plastic moment capacity. (,~) When properly designed and detailed, the extended cnd-platc connection can bc considered suitable for nloment-resisting frames in areas of high seismicity.

Acknowledgement The authors wish to acknowledge the support of the Natural Sciences and Engineering Research Council of

26

Eng. Struct. 1990, VoI. 12, J a n u a r y

Canada, This work is carried out under NSERC grants to McMaster University.

References 1

Kato, B.. Macda. Y. ;,nd Sakae. K. "Bchavk)ur of rigid frame suhds."ielllblltgc:.. :,,tlhiected to h~)rizontal force'. J,ints in Sl('clw,rk ("-,d:, reck,side Polytechnic. It)Xl

2 Popov, I-.P., Amin, N.R. and Stephen. R.M. "Cyclic behavior of large b e a m - c o l u m n ;.I..~sc+lnhlics steel connection>,'. I".ardutuakc .VpecIru 1985. l. No. 2. 203 23S 3 I'opov. [':.P. 'SeiF, ll|i+ l|lOll|+Clll COIIlleCIIoI1N foI- m o m e n t resisting steel frame>,'. EER(" Rt7,. 33 02. [ :niverxity t)f (:alifi)rnia. Berkely. Oct 19~3 4 Murray. T. M, "Recent ,.le,,el,.)pments for the design of moment endplate connections" J. ('¢m.strttcl. Ntccl Res. 19ig8, 10. 133 1¢)2 5 Zaxldonini, R. and /anon, I'. "l:xpcrinlcntal anal~,ds of cml plate ~tlllltC~tion:-,', Pro('. ,~'hHc-o/+lhc-~trt I I ' . r k s h . p . n ('~mnl,cli.ns, [.~tbortlt,lrc ~&" If('('¢ltllqltC t'l Jtt'('jtlftH/l~k'll'. { "lll+hilll. I"t++'l+'ll'U.I'lsevier Applied Science. 1~),';7.41 51 l~ John>,ton¢, N.l). and Walpole. Y~,.R "Beha+ior ol + steel beamO,',lUlTIn ¢'Jlln¢',.'lk)l|S. lll;idt~ ily, ill.~ bolted end plates', leull. ,%c)+ /cahtnd ,",at..",'o('./,r I:'arthqmJ,~c IEn~. 198". 15, No. 2. g2 92 7 Ballio. ( ; ,'t . I "Steel b e a m - t o - c o l u m n joints under c.~clic loads. experimental and analytical approach', l'r.c. ,~th I:'ur.pcan ('.nil ~)tt I;'arthquakc Fnt,,.. Lisbon. It}g6 ,x C a n a d i a n Standards Association. ('A 3,3-SI6, I I'rtqn+sall+w (+hm.+e 2(,. revi:.,cd gg 09 01. (._'~,nada ~) Americ:,n Institute of Steel (.+on.;tructi,.m. L R F D , .(~'peciltc~ltion/+~r ,~'lrttctttrtt/.~'lcc/ BuihlinL,~. (.'hicago. 19,';{~ IO Walpolc. W.R. and But,.'her. (LW. "P,earn design'. Bull. New

Extended end-plate connections under cyclic loading: A. Ghobarah et al. Zealand Nat. Soc. for Earthquake Eng. 1985, I& No. 4, I1 Canadian Institute of Steel Construction, Handbook of Steel Construction {4th edition), 1986 12 Grifl~ths, J.D. "End-plate moment connections-their use and misuse', Eng. Ji. AISC 1984, First Quarter, 32-34 13 Suttees, J.O. and Ibrahim, M.E. 'Bolt tension measurement from head strain data', J. Struct. Die., ASCE 1980, 106, No. ST?., 477--490 14 AImulti, A.M. and Hanson, R.D. "Static and dynamic cyclic yielding of steel beams" J. Struct. Div., ASCE 1973, 99, No. ST6, 1273-1285

15 Mann, A.P. and Morris, LJ. *Limit design of extended end-plate connections', J. Struct. Div., ASCE 19"/9, 105, No. ST3, 511-527 16 Packer, J.A. and Morris, LJ. 'A limit state design method for the tension region of bolted beam-column connections', The Struct. Eng. 1977, $, No. 10, 446-458 17 Krishnamurthy, N.A. *Fresh look at bolted end plate behavior and design', Eng. JI. AISC 1978, 2nd Quarter, 3949 18 Kulak, G.I., Fisher, J.W. and Struik, J.H. A Guide to Design Criteria for Bolted and Riveted Joints (2nd edition), John Wiley and Sons, 1987

Eng. Struct. 1990, Vol. 12, January 27