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Case study: Damage of an RC building after a landslide—inspection, analysis and retrofitting
Engineering Structures 32 (2010) 1814–1820
Contents lists available at ScienceDirect
Engineering Structures journal homepage: www.elsevier.com/locate/engstruct
Case study: Damage of an RC building after a landslide—inspection, analysis and retrofitting P. Tiago a,b , E. Júlio c,d,∗ a
EC+A – Projectos Lda, Coimbra, Portugal
b
Civil Engineering Department, Polytechnic Institute of Coimbra, Portugal
c
ISISE – Institute for Sustainability and Innovation in Structural Engineering, Portugal
d
Civil Engineering Department, University of Coimbra, Portugal
article
info
Article history: Available online 5 March 2010 Keywords: Damage Building Landslide Inspection Retrofitting Robustness
abstract In 2000, due to a substantial landslide, the reinforced concrete (RC) structure of a residential building located in Coimbra, Portugal, was severely damaged. The first two levels of three columns were completely destroyed and, as a result, part of the building supported by these, with a dimension in plant of 9.5 × 6.7 m2 , became a 7.0 m span cantilever with 12 stories. In this paper, the authors describe the following: the accident; the preliminary assessment of structural conditions; the immediate intervention; the strategy adopted to consolidate the damaged structure; the repair and strengthening works; the loading procedure of the rebuilt part of the structure; and the finishing operations. Some final remarks are also presented, including a proposal for robustness analysis. © 2010 Elsevier Ltd. All rights reserved.
1. Introduction In Portugal, the years between 1970 and 1995 were characterized by a considerably low quality of the construction sector. The main reasons contributing to this reality are the demographic migration from rural to urban areas that started in the beginning of this period and the 1974 revolution that gave rise to the present Portuguese democratic system. These two situations combined together led to an abnormal increase of construction associated to an equally peculiar reduction of quality standards in construction materials as well as in construction methods. At that time, in Portugal, residential buildings were generally designed and built adopting a structure of precast prestressed concrete beam and hollow clay block floors, with a cast-in-place concrete compressive layer, supported by reinforced concrete (RC) plane frames. Non-structural clay masonry walls were used both as partitioning walls and as external perimeter walls. In the latter situation, these were built as cavity walls, being therefore much thicker than the inner walls.
∗ Corresponding address: ISISE – Institute for Sustainability and Innovation in Structural Engineering, Faculty of Sciences and Technology, Department of Civil Engineering, Rua Luis Reis Santos(Polo II), 3030-788 Coimbra, Portugal. Tel.: +351 239797258; fax: +351 239797259. E-mail address: [email protected] (E. Júlio). 0141-0296/$ – see front matter © 2010 Elsevier Ltd. All rights reserved. doi:10.1016/j.engstruct.2010.02.018
As regards the structural design, it can be stated that Portuguese codes have always been of high quality, at least for concrete structures. This is due to an important research activity developed at the National Laboratory of Civil Engineering (LNEC) and also to the fact that Portuguese codes for reinforced concrete structures are generally based on CEB provisions. This paper describes the response of a residential building erected in the beginning of the 1980s in Coimbra, Portugal, subjected to an unforeseen event—the impact caused by a landslide, and describes how it was retrofitted. Lastly, some remarks are presented and a possible approach to enhance the robustness of this type of building is proposed. 2. Description of the accident The year 2000 was an unusually rainy year. Just three days before the new millennium, at 19:00 h, a substantial landslide occurred, causing severe damage in the RC structure of a 16-story residential building, erected in the beginning of the 1980s in Coimbra, Portugal (Fig. 1(a)). The first two levels of three columns were completely destroyed and, as a result, the rear body of the building supported by these, with a dimension in plant of 9.5 × 6.7 m2 , became a 7.0 m span cantilever with 12 stories (Fig. 1(b)). However, the damage could have been much more severe if the flow of that significant mass of soil and debris had not been damped by part of the 2-story parking garage located on the building’s backyard that completely vanished (Fig. 1(a)).
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Fig. 1. (a) Rear (West) façade of the building, a few days after the accident. (b) Detail of the total collapse of the outer columns of the damaged rear body of the building.
Fig. 2. Schematic drawing of the strut–tie system that materialized after the accident.
The building was evacuated a few hours after the accident and, at 12:00 h the next day, a visual inspection was carried out, mainly focused on the damaged rear body. On the outer masonry walls, no significant anomalies were identified. Inside, a few thin cracks distributed on the masonry walls were observed and larger cracks, with a maximum width of 2 mm, concentrated at the corners of the openings, were detected. 3. Preliminary assessment of the structural conditions The observed low level of damage was attributed to the joint behavior of the RC structure with the outer (non-structural) masonry walls that allowed a strut–tie system to materialize in order to resist the gravity loads. More specifically, in the damaged part of the building, the loads previously supported by the destroyed columns became equilibrated by compression stresses (struts) at the outer masonry walls and by tension stresses (ties) at the slabs (Fig. 2). In the remaining part of the building, this system originated: a tension resultant force at the top slab; a compression resultant force at the bottom slab; and an additional compression at the existing foundations (Fig. 2). In order to assess the safety of the damaged body, it was necessary to quantify the stress state in the resulting structural
system. This represented a major difficulty due to the existence of two sets of openings in both lateral walls. Therefore, and since a fast response was requested, it was decided to build a plane linear elastic finite elements model of these walls, including the columns common to the rest of the building. The RC structure was modeled using linear elements and the masonry walls were simulated with shell elements assuming adequate geometric and material properties. Namely, the North outer masonry walls, presenting larger openings, were assumed to be made of two 25 mm layers of mortar and a 70 mm layer of solid equivalent clay bricks, the latter corresponding to the effective width of the clay bricks, i.e., excluding the voids. To this element, with a total thickness of 120 mm, an equivalent Young’s modulus was attributed, based on an experimental study [1] conducted with similar mortar and clay bricks. Given the facts that (a) for mortar, a Young’s modulus of 3 GPa was found and, for clay bricks, the corresponding value was 10 GPa; (b) extrapolating these values from the mentioned research study [1] to the present case study comprises some uncertainty; and (c) it would be useful to evaluate the sensitiveness of results to this parameter, it was decided to conduct a parametric analysis, varying the equivalent Young’s modulus between 3 and 10 GPa. The effect of the remaining part of the building on the model was considered, assuming the nodes at the boundary to be horizontally restrained. From the 2D numerical analysis (Fig. 3(a)), it was possible to conclude that (1) the finite element approach validated the hypothesis of the strut–tie model behavior; (2) the stress state in the lateral masonry walls presented values ranging from −2 MPa to 2 MPa; (3) at the lower levels, concentration zones of principal tension stresses appeared at the corners of the openings with values up to 4 MPa; and (4) the prediction of the maximum deflection at the bottom level was between 3 and 8 mm. Since these results were in agreement with what was observed on site, this first approach was considered valid and the immediate intervention was planned based on it. Later, 3D finite element models were also developed and results (Fig. 3(b)) corroborated those of this first 2D modeling, although with some minor differences. The original project was also analyzed. It is relevant to mention that actions were quantified according to the 1961 RSEP [2], the Portuguese code on actions for buildings and bridges, and to the 1967 REBA [3], the Portuguese code on reinforced concrete structures, based on the 1963 CEB ‘‘Guidelines for Contractors’’ [4]. These codes already considered ultimate and service limit states as design criteria, instead of the traditional criteria, based on safety stresses. Nevertheless, the following relevant differences between these codes and the corresponding modern Eurocodes (EC), namely EC 0 [5], EC 1 [6], EC 2 [7] and EC 8 [8], can be identified in the scope of the present case study: (1) the combinations of actions
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Fig. 3. Stress results of the finite element analysis. (a) 2D model: Minimum principal stresses, on the left, and maximum principal stresses, on the right, both in MPa. (b) 3D model: Maximum principal stresses in MPa.
for seismic design situations according to RSEP [2] lead to lower values (approximately 50%) compared to EC 8 [8]; nevertheless, it should be mentioned that, in the original project, it is stated that this situation was considered to be more unfavorable than the one having the wind as the leading variable action; and (2) second-order effects are considered in REBA only locally, i.e., without considering the overall sway behavior of the structure [3]; thus columns would present higher reinforcement ratios, at least, if designed according to EC 2 [7]. Moreover, concerning the building’s erection, from the observation of damaged elements it was concluded that the connections were poorly accomplished and, although in the original project solid slabs were adopted, the contractor decided to replace these by prestressed concrete beam and hollow clay block floors. 4. Immediate intervention First of all, it should be stressed that the accident described had a major psychological impact on Coimbra’s inhabitants, in general, and on the building’s residents, in particular. Therefore, there was a considerable pressure from public opinion to have a fast and reliable retrofitting intervention. The 2D numerical analysis indicated that the structure was not at risk of imminent collapse. As mentioned above, at the outer masonry walls maximum tension stresses of approximately 4 MPa were located at the openings’ corners (where cracks appeared) and maximum compression stresses ranging between 3 and 5 MPa were identified. In the study [1] mentioned above, façade walls similar to the walls of the building analyzed here were tested in compression until failure, reaching ultimate values in the order of 13 MPa. Although the acting forces in these two cases did not have the same direction, given the significant differences in the corresponding values, the safety of the walls was assumed. Moreover, it was also assumed that the access to the damaged zones, corresponding to approximately more 15 kN per story (5% load increase), was still admissible concerning the walls. In relation to the slabs, given the need to have a fast solution and since the 3D model was being built, the safety check was based on simple calculations. According to these, the maximum tension resultant at the slabs was approximately 600 kN. Considering that
the ‘‘ties’’ (of the strut–tie system) were materialized by 1 m of slab width (thus by three precast prestressed concrete beams) below each outer wall combined with two peripheral RC beams, and taking into account a computed axial strength reserve of, respectively, 2 × 3 × 30 = 180 kN and 2 × 350 = 700 kN, totalizing 880 kN, the safety of the slabs was also assumed. Finally, regarding the foundations, it was estimated that with the accident the outer columns of the main body of the building became subjected to a 50% load increase. Nevertheless, according to the RC code effective when the building was designed, REBA [3], the characteristic values of both self-weight and imposed loads are multiplied by 1.5 in a fundamental combination of actions and variable action is considered in all stories. For this reason, following the accident, neither these columns nor the respective foundations were submitted to loads above the design values. Furthermore, the visual inspection conducted immediately after the accident did not reveal signs of settlements at the foundations or damage at the walls. Following the analysis described above, it was decided to undertake an immediate intervention from the inside, consisting in applying a provisional shoring system, aiming to reduce or at least to hold up the increase of the stress state of the masonry walls. This procedure also presented the advantage of being extremely simple. Extensible steel post shores were applied at some points in the interior and wooden bars were positioned in the openings (Fig. 4(a)). Time evolution of the cracks’ widths was registered at some locations, selected between those presenting a pattern more in agreement with the new structural system. As an example, Fig. 4(b) shows the crack at the corner of the window presented in Fig. 4(a). The corresponding daily record of the crack width was the following: 1.0 mm, at 11:30 h on December 29; 1.0 mm, at 14:00 h on December 30; and 1.0 mm, at 16:00 h on December 31. In this case, as in the remaining cases, it was observed that major cracks did not increase. In fact, only new and much smaller cracks could be noticed with time. For this reason, and also for security reasons, it was decided to stop monitoring cracks, just three days after the accident. At the outer walls, all cracks were painted (Fig. 5). Mapping the cracks had the major goal of facilitating the localization of these after the retrofitting operations, i.e., after closing them.
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Fig. 4. (a) Provisional shoring system applied inside. (b) Crack width monitoring during the first three days after the accident.
Fig. 5. Mapping cracks on the outer walls.
5. Strategy adopted for consolidation The immediate intervention referred to represented only a residual increase on the structure safety. Therefore, it was decided to proceed with a more substantial shoring organized in two phases. First, a self-equilibrated prestressing system was conceived and applied with the intention of consolidating the damaged body of the building, thus allowing the safe removal of debris in the accident zone. Then, with the area below the 12 stories being cantilever free, a second shoring system materialized with five steel shores was planned and applied. The prestressing system was intended to suspend the bottom part of the cantilever from the top level of the damaged rear
body using prestressing strands (Fig. 6(a)). This solution, besides its simplicity, presented the additional advantage of being selfequilibrated. In fact, it was relevant not to increase loads at the foundations since reliable data concerning the building footings was not available. A steel beam was placed on the top slab and two prestressing strands were anchored to it (Fig. 6(b)) with a prestress force of approximately 300 kN. This value is mainly justified by the urgency in applying the designed consolidating system, since the two S1600/1800 prestressing strands with a cross-section of 1.5 cm2 were adopted since they were available. Furthermore, the manufacturer recommended that the applied stress did not exceed 1000 MPa, thus giving 300 kN. It should be added that minimizing stresses on the top story’s slab was also a concern. For this reason, extensible steel post shores were also applied on both stories below, aiming to distribute the linear load between these three top stories. At both bottom corners of the cantilever, steel deviation devices, specifically designed for this end, were placed to change the strand direction (Fig. 6(c)). These devices were covered with a polytetrafluoroethylene (PTFE) layer to ensure that no prestress losses occurred during the application of the prestress force and also to avoid the introduction of lateral forces at these devices. With the prestressing system applied, the mass of soil and debris was removed from the accident zone and it was possible to execute the footings of the five steel shores and to put these in place (Fig. 7(a)). Afterwards, since an active shoring was wanted, a prestress force of approximately 150 kN was applied to each of these using hydraulic jacks (Fig. 7(b)). In this case, this value was chosen since 5 MPa was assumed as an acceptable upper limit for the stress increase locally introduced in the concrete, the loading area being 0.10 × 0.30 m2 . It should be stressed that the building was erected in a period characterized by a low quality level, as
Fig. 6. (a) Prestressing consolidation system. (b) Steel beam on top of the building and anchorage of prestressing tendons. (c) Steel deviation device.
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Fig. 7. (a) Provisional steel shores. (b) Prestressing the shores with hydraulic jacks.
Fig. 8. (a) Foundation for the retrofitting steel frame and shear wall. (b) Assembling the retrofitting steel frame. (c) Additional bracing steel shores.
mentioned before; the structure was damaged by the accident; and, according to the project, a B225 concrete (fck ≈ 18 MPa) was adopted. Lastly, a triangular steel bracing system was welded to the steel shores (Fig. 8(a)). 6. Repair and strengthening Once the consolidation operations of the damaged body of the building had been concluded, it was possible to move on with the rehabilitation work of both structural and non-structural elements. Since these operations were intended to take place as quickly as possible, and also aiming to minimize the timedependent deformations of members to rebuild, a composite steel and concrete solution was adopted. Instead of the original isolated footings, it was decided to adopt a combined footing for the three columns, aiming to obtain a uniform stress diagram at the soil/footing interface. Assuming a conservative design, a considerable contact area (11.5 × 3.0 m2 ) was adopted with the purpose of minimizing the stresses and consequently minimizing the time-dependent settlements. According to the geotechnical study referred to in the original project, later corroborated by an expert who collaborated in the retrofitting operation as a consultant, the building is founded on a hard sandstone soil, called Grés de Silves, typical in some regions of the Iberian peninsula. This soil presents an allowable pressure of 400 kPa but, for the reasons mentioned above, a design value of 200 kPa was assumed. The footing was executed with steel bolts on top to receive the retrofitting steel frame (Fig. 8(a)). The bottom rear masonry wall of the 12-story cantilever was replaced by an RC shear wall aiming to better accommodate the load transfer from both provisional shoring systems to the
retrofitting steel frame (Fig. 8(a)). Inside, near the columns, steel shores were also applied to strengthen these zones. Afterward, the retrofitting steel frame was assembled and connected to the foundation using grout to fill the voids (Fig. 8(b)). Provisional steel diagonals were welded to the retrofitting steel frame to serve as a bracing system during load transfer (Fig. 8(b)). In order to ensure an effective bracing also in the direction normal to the plane of the frame, three additional steel shores were also linked to the retrofitting steel frame (Fig. 8(c)). 7. Load transferring The connection between the retrofitting steel frame and the damaged body of the building was materialized by first applying an interface steel beam to the bottom of the latter with epoxybonded steel bolts and by filing the voids also with an epoxy resin. However, this operation could not be undertaken with the prestressing system installed. Therefore, first, the prestress force of each of the five steel shores of the second shoring system had to be slightly increased. Then, the prestressing strands and the deviation devices were removed. And, lastly, the interface steel beam was put in place. Provisional steel corbels were welded to the retrofitting steel frame on both sides of each column (Fig. 9). The aim was to transfer the load supported by the provisional shoring system to the retrofitting steel frame in one single step, using simultaneously six hydraulic jacks placed on top of these corbels (Fig. 9). Linking steel elements, specially designed to be placed between the interface steel beam (connected to the cantilever bottom) and the retrofitting steel frame, were then supposed to be welded to both
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Fig. 9. Schematic drawing of load transfer using hydraulic jacks on top of provisional steel corbels, welded to each of the retrofitting columns, and linking steel element to be placed between and welded to both the retrofitting column and the interface steel beam.
Fig. 10. Set up for load transfer using hydraulic jacks.
(Fig. 9). Nevertheless, since the contractor was only able to provide two hydraulic jacks, the load transfer was performed in several steps instead, using steel pads to be able to move the hydraulic jacks from one corbel to another and welding the three linking steel elements finally (Fig. 10).
a
b
Fig. 11. (a) Schematic drawing of the transfer steel plates welded to the interface steel beam and connected with M20 steel anchors to each of the existing RC columns. (b) Detail of the transfer steel plates.
8. Finishing operations After mobilizing the retrofitting steel frame, all steel shores from both the bracing system and the provisional shoring system were removed, as well as the steel corbels. Since, in the case of an earthquake, tension stresses between the retrofitting steel frame and the original structure will appear, it was decided to strengthen this connection. With this aim, transfer steel plates were welded to the interface steel beam and connected to each of the three RC columns with epoxy-bonded steel bolts (Fig. 11). The destroyed slab was rebuilt adopting a solid concrete slab supported by steel beams normal to the plane of the steel frame (Fig. 12(a)). In this process, steel elements from the provisional systems were reused. Next, all steel members were provided with steel reinforcement and covered with a high-performance grout (Fig. 12(b)) aiming to ensure effective fire and corrosion protection and also assumed as an additional strengthening measure. Finally, non-structural finishing operations were carried out and the building was painted (Fig. 12(b)). Later, the 2-story parking garage located in the backyard that had vanished was
Fig. 12. (a) Retrofitting steel frame after removing the provisional systems and rebuilding the destroyed slab. (b) Retrofitting operations concluded.
also rebuilt (Fig. 13) and the building retrofitting was completely accomplished.
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Fig. 13. Rebuilt parking garage.
9. Final remarks Relative to the intervention, it should be noted that the design of all structural members was performed considering the construction phases described. Consequently, the retrofitted structure turned out to be considerably more resistant than the original one. As an example, it can be mentioned that the load transfer phase constituted a more unfavorable loading situation, in terms of stresses and supporting conditions, than the final service conditions. Furthermore, it was ensured that the connection between the new and the existing structural members was compatible with the maximum strength capacity of these. As regards the original structure, it can be stated that it corresponds to a typical residential building erected in Portugal in the time period referred to in the introduction. In fact, the materials and methods at that time were not outstanding, for the reasons given, but, nevertheless, the building had a satisfactory response to the unforeseen event of an impact of a mass of soil and debris against its RC structure. For this reason, it can be stated that the structure of this building is robust, since it was capable of preventing a progressive collapse after a localized damage. This property, robustness, gained much interest after the Ronan Point block of flats disaster in 1968, caused by an explosion, and more recently after the Twin Towers catastrophe in 2001, caused by a terrorist attack. Different researchers have proposed different approaches to achieve structural robustness, including the creation of alternate
load paths in the structure; the improvement of redundancy and ductility; and the reduction of risk of abnormal loads, through protection of the structure [9]. Nevertheless, this concept is not yet included in codes, at least not in a helpful way. Taking the Eurocodes as an example, only in EC 1 [6] (part 7) does a definition of robustness appear: ‘‘the ability of a structure to withstand events like fire, explosions, impact or the consequences of human error, without being damaged to an extent disproportionate to the original cause’’, although no practical measures are provided. Some organizations have recently published some guidelines on robustness, such as IABSE [10], and there is a COST Action in progress to study this issue [11]. Nonetheless, common to all proposed approaches is the fact that these are always focused only on the structure. Following the experience described here, the authors would like to present a different suggestion concerning this subject. The building described in this paper was able to withstand the mentioned unforeseen event only because both lateral masonry walls were mobilized. It seems therefore rather interesting to start considering non-structural elements as a structural reserve, for this scenario only, not for fundamental, accidental or seismic combinations of actions. Naturally, specifications on the material properties and on the construction procedures, including detailing guidelines, of these non-structural elements will also be required. References [1] Vicente R. Pathology of facade walls—Mechanical behavior of facade walls with external correction of thermal bridges. M.Sc. thesis. University of Coimbra. 2002 [in Portuguese]. [2] RSEP. Code on actions for buildings and bridges. 1961 [in Portuguese]. [3] REBA. Code on reinforced concrete structures. 1967 [in Portuguese]. [4] Recommendations pratiques à l’usage des constructeurs. Comité Européen du Béton. 1963. [5] Eurocode 0: Basis of structural design. European Committee for Standardization. 2002. [6] Eurocode 1: Actions on structures. European Committee for Standardization. 2006. [7] Eurocode 2: Design of concrete structures. European Committee for Standardization. 2004. [8] Eurocode 8: Design of structures for earthquake resistance. European Committee for Standardization. 2004. [9] Ellingwood BR, Dusenberry DO. Building design for abnormal loads and progressive collapse. Comput Aid Civ Inf Eng 2005;20:194–205. [10] Knoll F, Vogel T. Design for robustness, structural engineering documents. Internat Assoc Bridge Struct Eng 2009. [11] COST-TU0601. Memorandum of understanding. 2007.
Contents lists available at ScienceDirect
Engineering Structures journal homepage: www.elsevier.com/locate/engstruct
Case study: Damage of an RC building after a landslide—inspection, analysis and retrofitting P. Tiago a,b , E. Júlio c,d,∗ a
EC+A – Projectos Lda, Coimbra, Portugal
b
Civil Engineering Department, Polytechnic Institute of Coimbra, Portugal
c
ISISE – Institute for Sustainability and Innovation in Structural Engineering, Portugal
d
Civil Engineering Department, University of Coimbra, Portugal
article
info
Article history: Available online 5 March 2010 Keywords: Damage Building Landslide Inspection Retrofitting Robustness
abstract In 2000, due to a substantial landslide, the reinforced concrete (RC) structure of a residential building located in Coimbra, Portugal, was severely damaged. The first two levels of three columns were completely destroyed and, as a result, part of the building supported by these, with a dimension in plant of 9.5 × 6.7 m2 , became a 7.0 m span cantilever with 12 stories. In this paper, the authors describe the following: the accident; the preliminary assessment of structural conditions; the immediate intervention; the strategy adopted to consolidate the damaged structure; the repair and strengthening works; the loading procedure of the rebuilt part of the structure; and the finishing operations. Some final remarks are also presented, including a proposal for robustness analysis. © 2010 Elsevier Ltd. All rights reserved.
1. Introduction In Portugal, the years between 1970 and 1995 were characterized by a considerably low quality of the construction sector. The main reasons contributing to this reality are the demographic migration from rural to urban areas that started in the beginning of this period and the 1974 revolution that gave rise to the present Portuguese democratic system. These two situations combined together led to an abnormal increase of construction associated to an equally peculiar reduction of quality standards in construction materials as well as in construction methods. At that time, in Portugal, residential buildings were generally designed and built adopting a structure of precast prestressed concrete beam and hollow clay block floors, with a cast-in-place concrete compressive layer, supported by reinforced concrete (RC) plane frames. Non-structural clay masonry walls were used both as partitioning walls and as external perimeter walls. In the latter situation, these were built as cavity walls, being therefore much thicker than the inner walls.
∗ Corresponding address: ISISE – Institute for Sustainability and Innovation in Structural Engineering, Faculty of Sciences and Technology, Department of Civil Engineering, Rua Luis Reis Santos(Polo II), 3030-788 Coimbra, Portugal. Tel.: +351 239797258; fax: +351 239797259. E-mail address: [email protected] (E. Júlio). 0141-0296/$ – see front matter © 2010 Elsevier Ltd. All rights reserved. doi:10.1016/j.engstruct.2010.02.018
As regards the structural design, it can be stated that Portuguese codes have always been of high quality, at least for concrete structures. This is due to an important research activity developed at the National Laboratory of Civil Engineering (LNEC) and also to the fact that Portuguese codes for reinforced concrete structures are generally based on CEB provisions. This paper describes the response of a residential building erected in the beginning of the 1980s in Coimbra, Portugal, subjected to an unforeseen event—the impact caused by a landslide, and describes how it was retrofitted. Lastly, some remarks are presented and a possible approach to enhance the robustness of this type of building is proposed. 2. Description of the accident The year 2000 was an unusually rainy year. Just three days before the new millennium, at 19:00 h, a substantial landslide occurred, causing severe damage in the RC structure of a 16-story residential building, erected in the beginning of the 1980s in Coimbra, Portugal (Fig. 1(a)). The first two levels of three columns were completely destroyed and, as a result, the rear body of the building supported by these, with a dimension in plant of 9.5 × 6.7 m2 , became a 7.0 m span cantilever with 12 stories (Fig. 1(b)). However, the damage could have been much more severe if the flow of that significant mass of soil and debris had not been damped by part of the 2-story parking garage located on the building’s backyard that completely vanished (Fig. 1(a)).
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Fig. 1. (a) Rear (West) façade of the building, a few days after the accident. (b) Detail of the total collapse of the outer columns of the damaged rear body of the building.
Fig. 2. Schematic drawing of the strut–tie system that materialized after the accident.
The building was evacuated a few hours after the accident and, at 12:00 h the next day, a visual inspection was carried out, mainly focused on the damaged rear body. On the outer masonry walls, no significant anomalies were identified. Inside, a few thin cracks distributed on the masonry walls were observed and larger cracks, with a maximum width of 2 mm, concentrated at the corners of the openings, were detected. 3. Preliminary assessment of the structural conditions The observed low level of damage was attributed to the joint behavior of the RC structure with the outer (non-structural) masonry walls that allowed a strut–tie system to materialize in order to resist the gravity loads. More specifically, in the damaged part of the building, the loads previously supported by the destroyed columns became equilibrated by compression stresses (struts) at the outer masonry walls and by tension stresses (ties) at the slabs (Fig. 2). In the remaining part of the building, this system originated: a tension resultant force at the top slab; a compression resultant force at the bottom slab; and an additional compression at the existing foundations (Fig. 2). In order to assess the safety of the damaged body, it was necessary to quantify the stress state in the resulting structural
system. This represented a major difficulty due to the existence of two sets of openings in both lateral walls. Therefore, and since a fast response was requested, it was decided to build a plane linear elastic finite elements model of these walls, including the columns common to the rest of the building. The RC structure was modeled using linear elements and the masonry walls were simulated with shell elements assuming adequate geometric and material properties. Namely, the North outer masonry walls, presenting larger openings, were assumed to be made of two 25 mm layers of mortar and a 70 mm layer of solid equivalent clay bricks, the latter corresponding to the effective width of the clay bricks, i.e., excluding the voids. To this element, with a total thickness of 120 mm, an equivalent Young’s modulus was attributed, based on an experimental study [1] conducted with similar mortar and clay bricks. Given the facts that (a) for mortar, a Young’s modulus of 3 GPa was found and, for clay bricks, the corresponding value was 10 GPa; (b) extrapolating these values from the mentioned research study [1] to the present case study comprises some uncertainty; and (c) it would be useful to evaluate the sensitiveness of results to this parameter, it was decided to conduct a parametric analysis, varying the equivalent Young’s modulus between 3 and 10 GPa. The effect of the remaining part of the building on the model was considered, assuming the nodes at the boundary to be horizontally restrained. From the 2D numerical analysis (Fig. 3(a)), it was possible to conclude that (1) the finite element approach validated the hypothesis of the strut–tie model behavior; (2) the stress state in the lateral masonry walls presented values ranging from −2 MPa to 2 MPa; (3) at the lower levels, concentration zones of principal tension stresses appeared at the corners of the openings with values up to 4 MPa; and (4) the prediction of the maximum deflection at the bottom level was between 3 and 8 mm. Since these results were in agreement with what was observed on site, this first approach was considered valid and the immediate intervention was planned based on it. Later, 3D finite element models were also developed and results (Fig. 3(b)) corroborated those of this first 2D modeling, although with some minor differences. The original project was also analyzed. It is relevant to mention that actions were quantified according to the 1961 RSEP [2], the Portuguese code on actions for buildings and bridges, and to the 1967 REBA [3], the Portuguese code on reinforced concrete structures, based on the 1963 CEB ‘‘Guidelines for Contractors’’ [4]. These codes already considered ultimate and service limit states as design criteria, instead of the traditional criteria, based on safety stresses. Nevertheless, the following relevant differences between these codes and the corresponding modern Eurocodes (EC), namely EC 0 [5], EC 1 [6], EC 2 [7] and EC 8 [8], can be identified in the scope of the present case study: (1) the combinations of actions
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Fig. 3. Stress results of the finite element analysis. (a) 2D model: Minimum principal stresses, on the left, and maximum principal stresses, on the right, both in MPa. (b) 3D model: Maximum principal stresses in MPa.
for seismic design situations according to RSEP [2] lead to lower values (approximately 50%) compared to EC 8 [8]; nevertheless, it should be mentioned that, in the original project, it is stated that this situation was considered to be more unfavorable than the one having the wind as the leading variable action; and (2) second-order effects are considered in REBA only locally, i.e., without considering the overall sway behavior of the structure [3]; thus columns would present higher reinforcement ratios, at least, if designed according to EC 2 [7]. Moreover, concerning the building’s erection, from the observation of damaged elements it was concluded that the connections were poorly accomplished and, although in the original project solid slabs were adopted, the contractor decided to replace these by prestressed concrete beam and hollow clay block floors. 4. Immediate intervention First of all, it should be stressed that the accident described had a major psychological impact on Coimbra’s inhabitants, in general, and on the building’s residents, in particular. Therefore, there was a considerable pressure from public opinion to have a fast and reliable retrofitting intervention. The 2D numerical analysis indicated that the structure was not at risk of imminent collapse. As mentioned above, at the outer masonry walls maximum tension stresses of approximately 4 MPa were located at the openings’ corners (where cracks appeared) and maximum compression stresses ranging between 3 and 5 MPa were identified. In the study [1] mentioned above, façade walls similar to the walls of the building analyzed here were tested in compression until failure, reaching ultimate values in the order of 13 MPa. Although the acting forces in these two cases did not have the same direction, given the significant differences in the corresponding values, the safety of the walls was assumed. Moreover, it was also assumed that the access to the damaged zones, corresponding to approximately more 15 kN per story (5% load increase), was still admissible concerning the walls. In relation to the slabs, given the need to have a fast solution and since the 3D model was being built, the safety check was based on simple calculations. According to these, the maximum tension resultant at the slabs was approximately 600 kN. Considering that
the ‘‘ties’’ (of the strut–tie system) were materialized by 1 m of slab width (thus by three precast prestressed concrete beams) below each outer wall combined with two peripheral RC beams, and taking into account a computed axial strength reserve of, respectively, 2 × 3 × 30 = 180 kN and 2 × 350 = 700 kN, totalizing 880 kN, the safety of the slabs was also assumed. Finally, regarding the foundations, it was estimated that with the accident the outer columns of the main body of the building became subjected to a 50% load increase. Nevertheless, according to the RC code effective when the building was designed, REBA [3], the characteristic values of both self-weight and imposed loads are multiplied by 1.5 in a fundamental combination of actions and variable action is considered in all stories. For this reason, following the accident, neither these columns nor the respective foundations were submitted to loads above the design values. Furthermore, the visual inspection conducted immediately after the accident did not reveal signs of settlements at the foundations or damage at the walls. Following the analysis described above, it was decided to undertake an immediate intervention from the inside, consisting in applying a provisional shoring system, aiming to reduce or at least to hold up the increase of the stress state of the masonry walls. This procedure also presented the advantage of being extremely simple. Extensible steel post shores were applied at some points in the interior and wooden bars were positioned in the openings (Fig. 4(a)). Time evolution of the cracks’ widths was registered at some locations, selected between those presenting a pattern more in agreement with the new structural system. As an example, Fig. 4(b) shows the crack at the corner of the window presented in Fig. 4(a). The corresponding daily record of the crack width was the following: 1.0 mm, at 11:30 h on December 29; 1.0 mm, at 14:00 h on December 30; and 1.0 mm, at 16:00 h on December 31. In this case, as in the remaining cases, it was observed that major cracks did not increase. In fact, only new and much smaller cracks could be noticed with time. For this reason, and also for security reasons, it was decided to stop monitoring cracks, just three days after the accident. At the outer walls, all cracks were painted (Fig. 5). Mapping the cracks had the major goal of facilitating the localization of these after the retrofitting operations, i.e., after closing them.
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Fig. 4. (a) Provisional shoring system applied inside. (b) Crack width monitoring during the first three days after the accident.
Fig. 5. Mapping cracks on the outer walls.
5. Strategy adopted for consolidation The immediate intervention referred to represented only a residual increase on the structure safety. Therefore, it was decided to proceed with a more substantial shoring organized in two phases. First, a self-equilibrated prestressing system was conceived and applied with the intention of consolidating the damaged body of the building, thus allowing the safe removal of debris in the accident zone. Then, with the area below the 12 stories being cantilever free, a second shoring system materialized with five steel shores was planned and applied. The prestressing system was intended to suspend the bottom part of the cantilever from the top level of the damaged rear
body using prestressing strands (Fig. 6(a)). This solution, besides its simplicity, presented the additional advantage of being selfequilibrated. In fact, it was relevant not to increase loads at the foundations since reliable data concerning the building footings was not available. A steel beam was placed on the top slab and two prestressing strands were anchored to it (Fig. 6(b)) with a prestress force of approximately 300 kN. This value is mainly justified by the urgency in applying the designed consolidating system, since the two S1600/1800 prestressing strands with a cross-section of 1.5 cm2 were adopted since they were available. Furthermore, the manufacturer recommended that the applied stress did not exceed 1000 MPa, thus giving 300 kN. It should be added that minimizing stresses on the top story’s slab was also a concern. For this reason, extensible steel post shores were also applied on both stories below, aiming to distribute the linear load between these three top stories. At both bottom corners of the cantilever, steel deviation devices, specifically designed for this end, were placed to change the strand direction (Fig. 6(c)). These devices were covered with a polytetrafluoroethylene (PTFE) layer to ensure that no prestress losses occurred during the application of the prestress force and also to avoid the introduction of lateral forces at these devices. With the prestressing system applied, the mass of soil and debris was removed from the accident zone and it was possible to execute the footings of the five steel shores and to put these in place (Fig. 7(a)). Afterwards, since an active shoring was wanted, a prestress force of approximately 150 kN was applied to each of these using hydraulic jacks (Fig. 7(b)). In this case, this value was chosen since 5 MPa was assumed as an acceptable upper limit for the stress increase locally introduced in the concrete, the loading area being 0.10 × 0.30 m2 . It should be stressed that the building was erected in a period characterized by a low quality level, as
Fig. 6. (a) Prestressing consolidation system. (b) Steel beam on top of the building and anchorage of prestressing tendons. (c) Steel deviation device.
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Fig. 7. (a) Provisional steel shores. (b) Prestressing the shores with hydraulic jacks.
Fig. 8. (a) Foundation for the retrofitting steel frame and shear wall. (b) Assembling the retrofitting steel frame. (c) Additional bracing steel shores.
mentioned before; the structure was damaged by the accident; and, according to the project, a B225 concrete (fck ≈ 18 MPa) was adopted. Lastly, a triangular steel bracing system was welded to the steel shores (Fig. 8(a)). 6. Repair and strengthening Once the consolidation operations of the damaged body of the building had been concluded, it was possible to move on with the rehabilitation work of both structural and non-structural elements. Since these operations were intended to take place as quickly as possible, and also aiming to minimize the timedependent deformations of members to rebuild, a composite steel and concrete solution was adopted. Instead of the original isolated footings, it was decided to adopt a combined footing for the three columns, aiming to obtain a uniform stress diagram at the soil/footing interface. Assuming a conservative design, a considerable contact area (11.5 × 3.0 m2 ) was adopted with the purpose of minimizing the stresses and consequently minimizing the time-dependent settlements. According to the geotechnical study referred to in the original project, later corroborated by an expert who collaborated in the retrofitting operation as a consultant, the building is founded on a hard sandstone soil, called Grés de Silves, typical in some regions of the Iberian peninsula. This soil presents an allowable pressure of 400 kPa but, for the reasons mentioned above, a design value of 200 kPa was assumed. The footing was executed with steel bolts on top to receive the retrofitting steel frame (Fig. 8(a)). The bottom rear masonry wall of the 12-story cantilever was replaced by an RC shear wall aiming to better accommodate the load transfer from both provisional shoring systems to the
retrofitting steel frame (Fig. 8(a)). Inside, near the columns, steel shores were also applied to strengthen these zones. Afterward, the retrofitting steel frame was assembled and connected to the foundation using grout to fill the voids (Fig. 8(b)). Provisional steel diagonals were welded to the retrofitting steel frame to serve as a bracing system during load transfer (Fig. 8(b)). In order to ensure an effective bracing also in the direction normal to the plane of the frame, three additional steel shores were also linked to the retrofitting steel frame (Fig. 8(c)). 7. Load transferring The connection between the retrofitting steel frame and the damaged body of the building was materialized by first applying an interface steel beam to the bottom of the latter with epoxybonded steel bolts and by filing the voids also with an epoxy resin. However, this operation could not be undertaken with the prestressing system installed. Therefore, first, the prestress force of each of the five steel shores of the second shoring system had to be slightly increased. Then, the prestressing strands and the deviation devices were removed. And, lastly, the interface steel beam was put in place. Provisional steel corbels were welded to the retrofitting steel frame on both sides of each column (Fig. 9). The aim was to transfer the load supported by the provisional shoring system to the retrofitting steel frame in one single step, using simultaneously six hydraulic jacks placed on top of these corbels (Fig. 9). Linking steel elements, specially designed to be placed between the interface steel beam (connected to the cantilever bottom) and the retrofitting steel frame, were then supposed to be welded to both
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Fig. 9. Schematic drawing of load transfer using hydraulic jacks on top of provisional steel corbels, welded to each of the retrofitting columns, and linking steel element to be placed between and welded to both the retrofitting column and the interface steel beam.
Fig. 10. Set up for load transfer using hydraulic jacks.
(Fig. 9). Nevertheless, since the contractor was only able to provide two hydraulic jacks, the load transfer was performed in several steps instead, using steel pads to be able to move the hydraulic jacks from one corbel to another and welding the three linking steel elements finally (Fig. 10).
a
b
Fig. 11. (a) Schematic drawing of the transfer steel plates welded to the interface steel beam and connected with M20 steel anchors to each of the existing RC columns. (b) Detail of the transfer steel plates.
8. Finishing operations After mobilizing the retrofitting steel frame, all steel shores from both the bracing system and the provisional shoring system were removed, as well as the steel corbels. Since, in the case of an earthquake, tension stresses between the retrofitting steel frame and the original structure will appear, it was decided to strengthen this connection. With this aim, transfer steel plates were welded to the interface steel beam and connected to each of the three RC columns with epoxy-bonded steel bolts (Fig. 11). The destroyed slab was rebuilt adopting a solid concrete slab supported by steel beams normal to the plane of the steel frame (Fig. 12(a)). In this process, steel elements from the provisional systems were reused. Next, all steel members were provided with steel reinforcement and covered with a high-performance grout (Fig. 12(b)) aiming to ensure effective fire and corrosion protection and also assumed as an additional strengthening measure. Finally, non-structural finishing operations were carried out and the building was painted (Fig. 12(b)). Later, the 2-story parking garage located in the backyard that had vanished was
Fig. 12. (a) Retrofitting steel frame after removing the provisional systems and rebuilding the destroyed slab. (b) Retrofitting operations concluded.
also rebuilt (Fig. 13) and the building retrofitting was completely accomplished.
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Fig. 13. Rebuilt parking garage.
9. Final remarks Relative to the intervention, it should be noted that the design of all structural members was performed considering the construction phases described. Consequently, the retrofitted structure turned out to be considerably more resistant than the original one. As an example, it can be mentioned that the load transfer phase constituted a more unfavorable loading situation, in terms of stresses and supporting conditions, than the final service conditions. Furthermore, it was ensured that the connection between the new and the existing structural members was compatible with the maximum strength capacity of these. As regards the original structure, it can be stated that it corresponds to a typical residential building erected in Portugal in the time period referred to in the introduction. In fact, the materials and methods at that time were not outstanding, for the reasons given, but, nevertheless, the building had a satisfactory response to the unforeseen event of an impact of a mass of soil and debris against its RC structure. For this reason, it can be stated that the structure of this building is robust, since it was capable of preventing a progressive collapse after a localized damage. This property, robustness, gained much interest after the Ronan Point block of flats disaster in 1968, caused by an explosion, and more recently after the Twin Towers catastrophe in 2001, caused by a terrorist attack. Different researchers have proposed different approaches to achieve structural robustness, including the creation of alternate
load paths in the structure; the improvement of redundancy and ductility; and the reduction of risk of abnormal loads, through protection of the structure [9]. Nevertheless, this concept is not yet included in codes, at least not in a helpful way. Taking the Eurocodes as an example, only in EC 1 [6] (part 7) does a definition of robustness appear: ‘‘the ability of a structure to withstand events like fire, explosions, impact or the consequences of human error, without being damaged to an extent disproportionate to the original cause’’, although no practical measures are provided. Some organizations have recently published some guidelines on robustness, such as IABSE [10], and there is a COST Action in progress to study this issue [11]. Nonetheless, common to all proposed approaches is the fact that these are always focused only on the structure. Following the experience described here, the authors would like to present a different suggestion concerning this subject. The building described in this paper was able to withstand the mentioned unforeseen event only because both lateral masonry walls were mobilized. It seems therefore rather interesting to start considering non-structural elements as a structural reserve, for this scenario only, not for fundamental, accidental or seismic combinations of actions. Naturally, specifications on the material properties and on the construction procedures, including detailing guidelines, of these non-structural elements will also be required. References [1] Vicente R. Pathology of facade walls—Mechanical behavior of facade walls with external correction of thermal bridges. M.Sc. thesis. University of Coimbra. 2002 [in Portuguese]. [2] RSEP. Code on actions for buildings and bridges. 1961 [in Portuguese]. [3] REBA. Code on reinforced concrete structures. 1967 [in Portuguese]. [4] Recommendations pratiques à l’usage des constructeurs. Comité Européen du Béton. 1963. [5] Eurocode 0: Basis of structural design. European Committee for Standardization. 2002. [6] Eurocode 1: Actions on structures. European Committee for Standardization. 2006. [7] Eurocode 2: Design of concrete structures. European Committee for Standardization. 2004. [8] Eurocode 8: Design of structures for earthquake resistance. European Committee for Standardization. 2004. [9] Ellingwood BR, Dusenberry DO. Building design for abnormal loads and progressive collapse. Comput Aid Civ Inf Eng 2005;20:194–205. [10] Knoll F, Vogel T. Design for robustness, structural engineering documents. Internat Assoc Bridge Struct Eng 2009. [11] COST-TU0601. Memorandum of understanding. 2007.