Cyclic loading behavior of EBF links constructed of ASTM A992 steel

Journal of Constructional Steel Research 63 (2007) 751–765 www.elsevier.com/locate/jcsr

Cyclic loading behavior of EBF links constructed of ASTM A992 steel Taichiro Okazaki a,∗ , Michael D. Engelhardt b a Department of Civil Engineering, University of Minnesota, Minneapolis, MN, 55455-0116, USA b Department of Civil, Architectural, and Environmental Engineering, University of Texas at Austin, Austin, TX 78712-0275, USA

Received 4 May 2006; accepted 8 August 2006

Abstract Cyclic loading tests were conducted to study the behavior of link beams in steel eccentrically braced frames. A total of thirty-seven link specimens were constructed from five different wide-flange sections, all of ASTM A992 steel, with link length varying from short shear yielding links to long flexure yielding links. The occurrence of web fracture in shear yielding link specimens led to further study on the cause of these fractures. Since the link web fracture appeared to be a phenomenon unique to modern rolled shapes, the potential role of material properties on these fractures is discussed. Based on the test data, a change in the flange slenderness limit is proposed. The link overstrength factor of 1.5, as assumed in the current U.S. code provisions, appears to be reasonable. The cyclic loading history used for testing was found to significantly affect link performance. Test observations also suggest new techniques for link stiffener design and detailing for link-to-column connections. c 2006 Elsevier Ltd. All rights reserved.

Keywords: Cyclic tests; Steel structures; Seismic design; Flange slenderness ratio; Loading history; Fracture; k-area; Eccentrically braced frame

1. Introduction The design intent for a seismic-resistant steel Eccentrically Braced Frame (EBF) is that inelastic action under strong earthquake motion is restricted primarily to the links. Therefore, the EBF design procedure prescribed in the 2005 AISC Seismic Provisions for Structural Steel Buildings [1] relies on an understanding of link behavior under severe cyclic loading. The AISC Seismic Provisions contain U.S. building code rules for detailing steel structures, including EBFs, for seismic resistance. The current building code rules for EBFs in the AISC Seismic Provisions, including link design, link rotation limits, and link overstrength factors, were developed from rather extensive experimental studies conducted almost exclusively on wide-flange shapes of ASTM A36 steel [2]. However, structural steel shapes most commonly used in the U.S. today are produced according to the newer ASTM A992 standard, which provides for a higher yield and tensile strength than A36 steel. The move to A992 steel raised concerns regarding the appropriateness of the flange width–thickness limits for EBF

∗ Corresponding author. Tel.: +1 612 626 0331; fax: +1 612 626 7750.

E-mail address: [email protected] (T. Okazaki). c 2006 Elsevier Ltd. All rights reserved. 0143-974X/$ - see front matter doi:10.1016/j.jcsr.2006.08.004

link sections. A limit of 0.30(E/Fy )1/2 was traditionally specified for the flange width–thickness ratio of EBF links. This flange slenderness ratio corresponds to 8.5 for A36 steel (with minimum specified yield strength of Fy = 250 MPa) and 7.2 for A992 steel (minimum specified Fy = 345 MPa). A number of rolled wide-flange shapes meet the flange slenderness limit of 8.5 but do not meet the limit of 7.2, and thus are disqualified from use as EBF links by the traditional flange slenderness limit. Meanwhile, the effect of flange slenderness ratio on link behavior has not been explicitly addressed in previous research. A secondary concern was the appropriateness of link overstrength factors used in the capacity design procedure for EBFs. Link overstrength is defined as the maximum shear force developed in the link divided by the plastic shear strength of the link. While the 2005 AISC Seismic Provisions implicitly assume a link overstrength factor of 1.5, recent tests on large built-up shear links for use in bridge applications showed overstrength factors of nearly 2 [3,4]. This has led to concerns that current overstrength factors may be unconservative, particularly for shapes with heavy flanges, where shear resistance of the flanges contributes significantly to overstrength. An experimental research program was conducted at the University of Texas at Austin. The initial objective for this program was to examine flange buckling and overstrength in

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Failure of the initial specimens to meet rotation requirements led to questions concerning the loading protocol for testing EBF links specified in the 2002 AISC Seismic Provisions [8]. A total of thirty-seven link specimens were tested during the course of this investigation. In the rest of this paper, the test specimens and test procedure is discussed, followed by an overview of test results. Key observations pertaining to loading history, link web fracture, and link end welds are discussed. The test data is used to evaluate the flange slenderness limit, inelastic rotation limit, and overstrength factor for EBF links prescribed in the 2005 AISC Seismic Provisions [1]. Suggestions on new techniques for link stiffener design and new detailing for link-to-column connections are made. While some results from the research program have been discussed in a previous publication by the authors [9], these results are also mentioned in this paper for completeness. 2. Experimental program 2.1. Test setup

Fig. 1. Test setup: (a) energy dissipation mechanism of EBF; (b) schematic representation of test setup; and (c) details and dimensions.

links constructed of A992 steel. Link specimens with various sections and lengths were tested for this purpose. As discussed later, many of these initial specimens failed prematurely due to fracture of the link web. This type of failure mode was not typically reported in earlier tests [5–7], and thus motivated further testing to investigate the cause of the link web fracture.

A test setup was devised to reproduce the force and deformation environment imposed on a link in an EBF with one end of the link attached to a column, as shown in Fig. 1(a). Fig. 1(b) illustrates that the kinematics of the test setup follows the energy dissipation mechanism of the EBF. Full details and dimensions of the test setup are shown in Fig. 1(c). The link length is indicated in the figures by the letter e. The link specimens were welded to heavy end plates at each end, as shown in Fig. 2. The end plates were, in turn, bolted into the setup, between the vertical column and horizontal beam.

Fig. 2. Details of selected link specimens.

T. Okazaki, M.D. Engelhardt / Journal of Constructional Steel Research 63 (2007) 751–765 Table 1 Test section properties Section W10×19 W10×33 W10×33 (A) W10×33 (B) W10×33 (C) W16×36 W10×68 W18×40

Fy (MPa) Flange Web

Fu (MPa) Flange Web

b f /2t f Nominal

Actual

367 356 374 379 374 362 319 352

509 507 520 518 518 534 479 499

5.1 9.1 9.1 9.1 9.1 8.1 6.6 5.7

5.2 9.2 9.6 9.2 9.6 7.1 6.6 6.1

405 382 365 402 367 392 404 393

531 503 503 530 503 565 531 527

Note: The tabulated Fy is a static yield stress value, measured with the test machine cross-heads stationary. The tabulated Fu is a dynamic ultimate strength, measured with the test machine cross-heads in motion.

2.2. Test specimens Five different wide-flange shapes were used to construct the test specimens. All sections were of ASTM A992 steel. The actual measured yield and ultimate strength values are listed in Table 1 for samples taken from the edges of the flanges and from mid-depth of the web. The link sections had a range of flange width–thickness ratios, b f /2t f , to study the effect of flange slenderness on link behavior. The W10×19, W10×68, and W18×40 sections satisfied the seismically compact limit [1] for flanges of 0.30(E/Fy )1/2 (=7.2 for Fy = 345 MPa). The W10×33 was chosen specifically because its flange slenderness exceeds the seismically compact limit and is at the compact limit [10] of 0.38(E/Fy )1/2 (=9.2 for Fy = 345 MPa). The flange slenderness of the W16×36, based on the actual measured dimensions, was substantially smaller than the nominal slenderness so that this section satisfied the seismically compact limit. Another consideration for selecting the test sections was to study the link overstrength for sections with large ratios of flange to web area. This was based on a concern that heavy flanges can contribute substantially to the shear capacity of the section, and therefore generate high levels of overstrength. The W10×68 section was chosen specifically to investigate this issue. The ratio of the area of one flange to the area of the web, for the W10×68, is approximately 2, which is near an upper bound for rolled wide-flange shapes normally used as links. Table 2 provides a listing of all link test specimens. A range of link lengths were tested, from short shear yielding links to long flexural yielding links. Links with a length less than 1.6M p /V p are dominated by shear yielding, whereas those longer than 2.6M p /V p are dominated by flexural yielding [1]. Between these limits, link inelastic response is heavily influenced by both shear and flexure. The link length parameter, e/(M p /V p ), listed in Table 2, was evaluated based on the measured section dimensions and the measured yield strength values. Fig. 2 shows schematic views of specimens with various stiffener details. The stiffener locations are listed in Table 2. As indicated in the table, the thirty-seven link specimens are categorized into three groups. All specimens in Groups I and II were provided with intermediate stiffeners according to the

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2005 AISC Seismic Provisions. As illustrated in Fig. 2 for Specimens 4A, 4B, and 4C, stiffeners were provided on only one side of the web, as permitted by the Provisions for link sections with a depth less than 635 mm. The stiffeners were full depth, welded to the web and to both flanges using fillet welds. The specimens in Group III, which were all identical in section and length, were provided with varying stiffener details. Specimens S1 through S3 had full depth one-sided stiffeners, similar to specimens in Groups I and II. Specimens S4, S6, and S9 had full depth stiffeners at both sides of the web, welded to the web and to both flanges. Specimens S5, S7, S8, and S10 did not meet the stiffener requirements in the Provisions. Specimens S5 and S8 had stiffeners at both sides of the web, welded only to the flanges and not to the web. Specimen S7 had one-sided stiffeners welded only to the web and not to the flanges. The spacing of stiffeners in Specimen S10 did not meet the stiffener spacing requirement in the Provisions. The stiffeners for Specimens S6 and S9 were welded to the web and flanges using a self-shielded flux core arc welding (FCAW) process with an E70T-6 electrode. In the remaining thirty-five specimens, the stiffeners were welded to the web and/or flanges using a shielded metal arc welding (SMAW) process with an E7018 electrode. 2.3. Loading protocol Four different cyclic loading protocols, as shown in Fig. 3, were used in the tests. As indicated in the figure, the four protocols are referred to in this paper as the old-AISC, revised, severe, and random loading protocols. Each loading protocol controls the link rotation angle, γ , which is computed as the relative displacement of one end of the link compared to the other, divided by the link length. After several initial elastic cycles, the old-AISC loading protocol (Old) requires increasing the applied link rotation in increments of 0.01 rad, with two cycles of loading applied at each increment of rotation. The severe loading protocol (SEV) was identical to the old-AISC protocol, except that four cycles of loading, instead of two cycles, were required at each increment of rotation. This protocol was intended to promote low cycle fatigue and premature failure of the link specimen. The Revised Loading Protocol (RLP) requires that, after completing the loading cycle at a link rotation of 0.05 rad, the link rotation be increased in increments of 0.02 rad, with one cycle of loading applied at each increment of rotation. The old-AISC protocol was specified in the previous, 2002 AISC Seismic Provisions as the loading protocol for testing EBF links. However, the old-AISC protocol is replaced by the revised protocol in the current, 2005 AISC Seismic Provisions. As discussed later, the revised protocol is believed to be more representative of demands caused by actual earthquake ground motion than the old-AISC protocol. Except for Specimen 9-RLP, the revised protocol used in this study included two loading cycles at a link rotation of 0.02 rad, instead of four loading cycles required in the 2005 AISC Seismic Provisions. It is believed that the lack of these small amplitude cycles had limited influence on the overall performance of the specimen.

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Table 2 Test specimens Group

Specimen

Section

Link length e (mm)

e/(M p /V p )

Intermediate stiffeners

Loading protocol

I

1A 1B 1C 2 3 4A 4B 4C 5 6A 6B 7 8 9 10 11 12

W10×19 W10×19 W10×19 W10×19 W10×19 W10×33 W10×33 W10×33 W10×33 W10×33 W10×33 W10×33 W16×36 W16×36 W10×68 W10×68 W18×40

584 584 584 762 1219 584 584 584 930 1219 1219 1854 930 1219 930 1219 584

1.73 1.73 1.73 2.25 3.61 1.04 1.04 1.04 1.65 2.16 2.16 3.29 1.49 1.95 1.25 1.64 1.02

3@146 mm 3@146 mm 3@146 mm 4@152 mm 152 mm from each end 3@146 mma 3@146 mma 3@146 mma 5@156 mm 4@244 mm 4@244 mm 305 mm from each end 6@133 mm 5@203 mm 2@305 mm 3@305 mm 3@146 mm

Old Old Old Old Old Old Old Old Old Old Old Old Old Old Old Old Old

II

4A-RLP 4C-RLP 8-RLP 9-RLP 10-RLP 11-RLP 12-RLP 12-MON 12-SEV 12-RAN

W10×33 W10×33 W16×36 W16×36 W10×68 W10×68 W18×40 W18×40 W18×40 W18×40

584 584 930 1219 930 1219 584 584 584 584

1.04 1.04 1.49 1.95 1.25 1.64 1.02 1.02 1.02 1.02

3@146 mma 3@146 mma 6@133 mm 5@203 mm 2@305 mm 3@305 mm 3@146 mm 3@146 mm 3@146 mm 3@146 mm

RLP RLP RLP RLP RLP RLP RLP MON SEV RAN

III

S1 S2 S3 S4 S5 S6 S7 S8 S9 S10

W10×33 (A) W10×33 (B) W10×33 (C) W10×33 (B) W10×33 (B) W10×33 (B) W10×33 (B) W10×33 (B) W10×33 (B) W10×33 (B)

584 584 584 584 584 584 584 584 584 584

1.01 0.99 0.99 0.99 0.99 0.99 0.99 0.99 0.99 0.99

3@146 mma 3@146 mma 3@146 mma 3@146 mma 3@146 mma,b 3@146 mma 3@146 mma,b 3@146 mma,b 3@146 mma 2@195 mma,b

SEV SEV SEV SEV SEV SEV SEV RLP RLP RLP

a See Fig. 2 for stiffener details. b Violates the stiffener requirements in the 2005 AISC Seismic Provisions.

Finally, the random loading protocol (RAN) was a randomly generated sequence which imposes large rotations in both loading directions during early loading cycles. As indicated in Table 2, the seventeen specimens in Group I were tested using the old-AISC protocol. The ten specimens in Group II were tested using various loading histories, including monotonic loading (MON). The ten specimens in Group III used either the severe protocol or the revised protocol. 3. Test results Acceptance criteria for links are defined in the 2005 AISC Seismic Provisions based on inelastic rotation. The inelastic rotation, γ p , is evaluated by removing the contributions of elastic response from the link rotation, γ . The Provisions specify shear yielding links (e ≤ 1.6M p /V p ) should be capable of developing an inelastic rotation of 0.08 rad, whereas flexural yielding links (e ≥ 2.6M p /V p ) should be capable of an

inelastic rotation of 0.02 rad. The required inelastic rotation of intermediate length links (1.6M p /V p < e < 2.6M p /V p ) is determined by linear interpolation between 0.08 and 0.02 rad. The inelastic rotation capacity of the link specimens was defined per the 2005 AISC Seismic Provisions, as the maximum level of inelastic rotation sustained for at least one full cycle of loading prior to the link shear strength dropping below the nominal link shear strength. Here, the nominal strength was evaluated based on the nominal section dimensions and a yield strength of Fy = 345 MPa. Table 3 summarizes results for each of the link specimens tested in this program. The table lists the actual inelastic rotation achieved by the specimen, along with the inelastic rotation required by the 2005 AISC Seismic Provisions. Also listed is a brief description of the controlling failure mode for each specimen. Detailed descriptions of individual tests can be found in Arce [11] (for all specimens in Group I except Specimen 12), Ryu [12] (Specimen 12 and all specimens

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Fig. 3. Loading protocols.

Fig. 4. Response of Specimen 4A.

in Group II except Specimen 9-RLP), and Galvez [13] (all specimens in Group III). Specimens 1A, 1B, and 6A failed prematurely due to fractures at the fillet welds connecting the link flanges to the end plates. These failures are considered an artifact of the test setup, as the link end connections used for the specimens were not representative of typical link end connection details used in actual EBFs. Details of the fracture in the link end welds and their implications will be discussed later in this paper. Meanwhile, these specimens will be excluded from the discussion of link behavior. Excluding Specimens 1A, 1B, and 6A, there are thirty-four remaining specimens that were not affected by failures at the link end connections, and can therefore be considered as providing valid fundamental data on the behavior of links.

A notable feature of the tests was that specimens with length e < 1.7M p /V p typically exhibited link web fracture as the controlling failure mode. The link web fractures initiated at the top and bottom ends of the link web stiffeners, at the point of termination of the fillet welds connecting the stiffeners to the link web. These fractures often propagated in a horizontal direction, running parallel to the flanges. Ultimately, the growth of these cracks led to a drastic reduction of the link shear resistance. Specimen 4A provides an example of a specimen that failed due to this type of fracture. Fig. 4 shows the hysteretic response of this specimen and Fig. 5 shows the specimen after testing. Fractures running across the top and bottom of the link web are visible in this photo. The link web fractures limited the inelastic rotation capacity of Specimen 4A to γ p = 0.06 rad, falling short of the 0.08 rad required by the

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Table 3 Test results Specimen 1A 1B 1C 2 3 4A 4B 4C 5 6A 6B 7 8 9 10 11 12 4A-RLP 4C-RLP 8-RLP 9-RLP 10-RLP 11-RLP 12-RLP 12-MON 12-SEV 12-RAN S1 S2 S3 S4 S5 S6 S7 S8 S9 S10

γ p (rad) Required

Test

0.072 0.072 0.072 0.042 0.02 0.08 0.08 0.08 0.077 0.046 0.046 0.02 0.08 0.059 0.08 0.078 0.08 0.08 0.08 0.08 0.059 0.08 0.078 0.08 0.08 0.08 0.08 0.08 0.08 0.08 0.08 0.08 0.08 0.08 0.08 0.08 0.08

0.042 0.060 0.081 0.070 0.041 0.061 0.071 0.080 0.067 0.047 0.047 0.037 0.077 0.048 0.073 0.068 0.091 0.12 0.12 0.117 0.058 0.113 0.087 0.119 >0.34 0.072 0.125 0.062 0.061 0.072 0.061 0.071 0.051 0.051 0.122 0.101 0.121

Observed failure mode Fracture at link end plate connection Fracture at link end plate connection Flange and web buckling followed by fracture in web Flange and web buckling followed by fracture in flange near link end Flange and web buckling followed by fracture in flange near link end Fracture of web at stiffener weld Fracture of web at stiffener weld Fracture of web at stiffener weld Fracture of web at stiffener weld Fracture at link end plate connection Flange and web buckling followed by fracture in web Flange, web, and lateral torsional buckling Flange and web buckling followed by fracture in web at stiffener weld Flange and web buckling Fracture of web at stiffener weld Fracture of web at stiffener weld Fracture of web at stiffener weld accompanied by web buckling Fracture of web at stiffener weld Fracture of web at stiffener weld Flange and web buckling followed by fracture in web at stiffener weld Flange and web buckling Fracture of web at stiffener weld possibly caused by inadequately small fillet weld size Fracture of web at stiffener weld; link rotation limited by ram stroke Fracture of web at stiffener weld accompanied by web buckling Web buckling Fracture of web at stiffener weld Web buckling followed by fracture of web at stiffener weld Fracture of web at stiffener weld Fracture of web at stiffener weld Fracture of web at stiffener weld Fracture of web at stiffener weld Web buckling followed by fracture of web Fracture of web at stiffener weld Fracture of web at stiffener weld promoted by web buckling Web buckling followed by fracture of web Fracture of web at stiffener weld Fracture of web at stiffener weld and web buckling

2005 AISC Seismic Provisions. In fact, the majority of shear link specimens in Group I, tested with the old-AISC loading protocol, failed to achieve their required inelastic link rotations due to this type of fracture. The majority of the twenty-seven specimens with length e < 1.7M p /V p were controlled by link web fracture. The only exceptions were Specimen 12-MON, which was tested with monotonic loading, and Specimens S5 and S8, which did not have the stiffeners welded to the link web. Specimens 9 and 9-RLP were identical specimens with an intermediate length of e = 2.0M p /V p , tested with the old-AISC protocol and revised protocol, respectively. Both specimens failed before achieving the rotation requirement, due to strength degradation associated with severe flange and web buckling in the end panels. The link web fractures discussed above were not observed in these specimens. Fig. 6 shows the hysteretic response of Specimen 9-RLP and Fig. 7 shows the specimen after testing. The photo shows substantial concentration of deformation in the end panels, where the section was severely distorted due to combined flange and web

Fig. 5. Specimen 4A after testing.

buckling. Significant yielding is visible in the link web panels besides the end panels. The development of local buckling led to the gradual strength degradation shown in Fig. 6. Besides Specimens 9 and 9-RLP, six other specimens had longer links in the intermediate and flexure yielding range of

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Fig. 6. Response of Specimen 9-RLP.

Fig. 7. Specimen 9-RLP after testing.

1.7M p /V p < e ≤ 3.6M p /V p . These specimens, all tested with the old-AISC loading protocol, successfully achieved their required link rotations, and ultimately failed due to combinations of severe flange buckling, web buckling, and in some cases, lateral torsional bucking. The link web fracture exhibited by short links was not observed in these specimens. Overall, the behavior of longer links (1.7M p /V p < e ≤ 3.6M p /V p ) was very similar to that reported in earlier tests, for example in Engelhardt and Popov [14]. 4. Loading protocol A large number of test specimens in Group I failed prematurely, before achieving their required rotation levels. Most of these specimens met all link design requirements, and were tested with the old-AISC loading protocol specified in the 2002 AISC Seismic Provisions. After observing these results, Richards and Uang [15] noted that the typical loading histories used in shear link tests conducted in the 1970s and 1980s introduced a significantly smaller number of inelastic loading cycles compared to the old-AISC loading protocol used for the tests in Group I. Further, there appeared to be no rational basis for the old-AISC loading protocol. Consequently, Richards and Uang [16,17] developed a revised loading protocol for testing

EBF links, which was then used for selected specimens in Groups II and III. The revised loading protocol was developed using a methodology similar to that used for moment frame connection testing, developed under the FEMA/SAC program by Krawinkler et al. [18]. Seven of the specimens in Group I that failed to meet their inelastic rotation requirements were duplicated and retested using the revised loading protocol developed by Richards and Uang [16,17]. Among the seven specimens, the six shorter specimens with length e < 1.7M p /V p achieved link rotations well in excess of the required level. As indicated by the data in Table 3, these specimens developed inelastic rotations of 10–50% greater than the level required in the 2005 AISC Seismic Provisions. The inelastic rotation developed by Specimen 11-RLP was limited due to limitations in the stroke of the loading ram. Nonetheless, this specimen developed an inelastic rotation exceeding the required level. However, one specimen tested with the revised loading protocol, 9-RLP, failed to significantly exceed the required rotation. After completing a full cycle at γ p = ±0.058 rad, this intermediate link specimen lost its strength during the following loading cycle of roughly γ p = ±0.08 rad due to severe flange and web buckling. Consequently, although Specimen 9-RLP achieved a 20% greater rotation compared to Specimen 9, which was tested with the old-AISC loading protocol, Specimen 9-RLP still failed to meet the required inelastic rotation of γ p = 0.059 rad. However, the specimen retained a shear strength of 120% of its nominal shear strength at γ p = ±0.058 rad, and the following strength degradation was rather gradual. As indicated in Fig. 6, the backbone curve connecting the maximum rotation points at each rotation increment exceed the nominal shear strength up to γ p = ±0.07 rad, which is beyond the required γ p = 0.059 rad. Therefore, since Specimen 9-RLP failed to meet the rigorous inelastic rotation requirement by such a small margin, and maintained appreciable strength during rotations beyond the required inelastic rotation, this specimen is considered as effectively satisfying the rotation requirements of the 2005 AISC Seismic Provisions.

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Fig. 8. Response of Specimen 4A-RLP.

Fig. 9. Specimen 4A-RLP after testing.

To illustrate the effect of loading protocol, Fig. 4 shows the hysteretic response of Specimen 4A, which was tested using the old-AISC loading protocol. In comparison, Fig. 8 shows the hysteretic response of Specimen 4A-RLP, which was nominally identical to Specimen 4A, except that 4A-RLP was tested using the revised loading protocol. Specimen 4A-RLP developed an inelastic rotation capacity of γ p = 0.12 rad, as compared to γ p = 0.061 rad for Specimen 4A. Fig. 9 is a photo of Specimen 4A-RLP after testing, showing the link web fractures that ultimately caused failure of this specimen. This failure mode is very similar to that observed in Specimen 4A, shown in Fig. 5. In general, whereas the change from the old-AISC protocol to the revised protocol led to an increase in link rotation capacity on the order of 20–100%, it did not significantly change the controlling failure mode for the link. Links that failed due to fracture of the link web under the oldAISC protocol still failed by fracture of the link web under the revised protocol. In order to further investigate the effect of loading history on link behavior, four replicates of Specimen 12 were fabricated and tested with different loading histories. In addition to the revised loading protocol discussed above (Specimen 12-RLP), three other duplicate specimens were subjected to monotonic loading (12-MON), the severe loading protocol (12-SEV),

and random loading protocol (12-RAN). Specimen 12-MON achieved an inelastic rotation larger than 0.34 rad, which is more than four times the 0.08 rad required in the 2005 AISC Seismic Provisions. An earlier monotonic loading test by Kasai and Popov [7] also showed a shear link specimen developing γ p = 0.19 rad. These test results demonstrate the ability of EBF links to withstand very large single excursion deformations, such as those imposed by large near-fault earthquake ground motions. The inelastic rotations achieved in the five tests on the replicates of Specimen 12 were strongly related to the imposed loading history. The inelastic rotation capacities increased from Specimen 12-SEV (γ p = 0.072 rad), to 12 (γ p = 0.091 rad), to 12-RLP (γ p = 0.119 rad), to 12-RAN (γ p = 0.125 rad), and finally, to 12-MON (γ p > 0.34 rad, one direction only). Specimens 12, 12-RLP, 12-SEV, and 12-RAN all failed due to link web fracture initiating at the termination of stiffener to link web welds. Specimen 12-MON failed due to severe web buckling and distortion of the link. Although web fracture was observed at the stiffener weld terminations, web buckling was the primary cause of strength loss. Specimen 12-RAN showed strength degradation due to web buckling before development of significant link web fracture. These observations suggest that link web fracture is related to low cycle fatigue effects, and if very large rotations are imposed at an early loading stage, significant web buckling can precede the occurrence of link web fracture. As illustrated by the discussion above, the loading protocol used to test link specimens has a very large effect on the inelastic rotation capacity achieved by the link. Links tested with the revised loading protocol achieved inelastic rotations that were from 20 to 100% greater than links tested with the old-AISC protocol. The average increase in inelastic rotation using the revised protocol over using the old-AISC protocol was 47%. Since the loading protocol has such a large effect on link test results, it is important that loading protocols be selected that realistically reflect link demands under actual earthquake loading, as represented by the revised loading protocol developed by Richards and Uang [16,17].

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5. Link web fracture The majority of specimens in the length range of e < 1.7M p /V p developed web fractures at the ends of stiffener to link web welds, prior to the occurrence of any web buckling, or only after very mild buckling, as exhibited by Specimen 4A (see Fig. 5). However, this type of link web fracture was not reported in earlier tests, for example in Hjelmstad and Popov [5], and Malley and Popov [6]. While fractures of the link web were reported in these past tests, those fractures occurred only after the web had undergone severe buckling, with fracture initiation occurring at locations of large localized bucking deformations. An exception is a recent test by McDaniel et al. [4] on a very large built-up link section tested for use in the new east span of the San Francisco–Oakland Bay Bridge. In this test, a fracture initiated at the termination of a stiffener weld prior to web buckling, although the fracture propagated diagonally across the web, compared to the horizontal fracture propagation observed in Specimen 4A. Analysis of the failure [4] suggested that the fracture was caused by a stress concentration at the end of the stiffener weld, because the stiffener was terminated too close to the flange-to-web groove weld of the built-up section. The occurrence of pre-bucking web fracture as the controlling failure mode for shear links was not typically observed in the extensive link testing programs conducted in the 1980s, which formed the basis for EBF detailing rules in the AISC Seismic Provisions. Consequently, links tested in this current program exhibited fundamentally different failure mechanisms from many links tested in earlier programs. Richards and Uang [15] noted three significant differences between recent and earlier link tests. These differences were in: (a) cyclic loading history; (b) stiffener details; and (c) link material. Tests in Groups II and III were conducted to further study the effect of these three factors, and to further examine the cause of the link web fracture. 5.1. Loading protocol An initial concern after completing the tests in Group I was that the majority of shear yielding links failed due to link web fracture before achieving the required inelastic rotation of γ p = 0.08 rad. As discussed earlier, this concern led to questions regarding the loading protocol provided in the 2002 AISC Seismic Provisions, and the development of the revised loading protocol by Richards and Uang [16,17]. Six shear yielding links were duplicated and retested using the revised loading protocol. Since all six specimens (4A-RLP, 4C-RLP, 8-RLP, 10-RLP, 11-RLP, and 12-RLP) significantly exceeded the required γ p = 0.08 rad, concerns regarding the ability of shear links to provide adequate inelastic rotation capacity were largely alleviated. The shear yielding links tested in this program consistently failed by link web fracture, regardless of the loading protocol. The only exceptions were Specimens 12-MON, S5, and S8. While the monotonically loaded Specimen 12-MON failed due to web buckling, the web buckling was not evident until the specimen reached a very large rotation of γ p = 0.2 rad. The

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failure modes of Specimens S5 and S8 were associated with the unique stiffener details used for these specimens, in that the stiffeners were not welded to the link web. These unique specimens aside, all other test results indicate that the loading history has little influence on the link web fracture. That is, regardless of loading protocol, the shear links tested in this program consistently exhibited web fracture as the controlling failure mode. 5.2. Stiffener details The first specimen in this program to exhibit a horizontal web fracture initiating at the end of a stiffener weld was Specimen 4A. Following this failure, and based on the recommendations by McDaniel et al. [4], the stiffener welds were terminated at a larger distance from the flange in the subsequent Specimens 4B and 4C. In going from Specimens 4A to 4B, and then to 4C, the termination of the stiffener weld was moved progressively further from the flange (see Fig. 2). In Specimen 4C, the stiffener welds were terminated a distance of approximately five times the web thickness from the “k-line” of the section. The k-line is the location where the web meets the flange–web fillet. The test results (see Table 3) show that larger inelastic rotations were achieved as the stiffener welds were moved further from the k-line. However, even for Specimen 4C, which had the stiffener welds terminated quite a large distance from the k-line, the horizontal fractures still ultimately developed. Thus, while moving the stiffener welds further from the k-line was beneficial, it did not eliminate the occurrence of link web fracture. Although moving the stiffener weld termination further from the k-line resulted in higher inelastic rotations prior to link web fracture, all shear link specimens tested with the revised loading protocol developed the required inelastic rotation of 0.08 rad. That is, even specimens where the stiffener weld termination was relatively close to the k-line (Specimen 4ARLP for example) still developed 0.08 rad inelastic rotation. Therefore, providing a generous distance between the k-line and the stiffener weld termination, while beneficial, may not be essential for satisfactory link performance. Nonetheless, due to the beneficial higher inelastic rotations, terminating the stiffener welds a generous distance from the k-line is recommended. Based on this test program, a distance of five times the link web thickness from the k-line of the link section to the stiffener weld termination is suggested as a reasonable basis for design. As noted by Richards and Uang [15], a large number of earlier link tests were conducted to investigate stiffener design criteria, and as such, the majority of these specimens did not satisfy the current stiffener design criteria. As discussed above, the link web fracture observed in Groups I and II tests was not reported previously with the exception of McDaniel et al. [4], which also followed the current stiffener design criteria. Therefore, it was speculated that the preclusion of web buckling shifts the critical failure mode to one controlled by fracture at the location of high restraints due to low cycle fatigue. In order to investigate the relation between stiffener

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details and link web fracture, and to identify details that delay link web fracture, specimens in Group III were tested with various stiffener details. Specimens S4 through S10 were provided with non-typical stiffener details, some of which violate requirements in the 2005 AISC Seismic Provisions. These specimens were constructed of the same material as Specimen S2, which conformed to the stiffener requirements in the Provisions. The severe loading protocol was used for Specimens S4 through S7 in order to promote premature link web fracture. Further, the specimens had stiffener welds terminated close to the k-line of the section (see Fig. 2) in order to draw higher stresses and strains near the k-area, where rolled shapes can have locally degraded material properties. Specimen S4 had full depth stiffeners on both sides of the web, welded to both flanges and to the web. This arrangement was expected to elevate the local restraint near the k-area compared to the standard arrangement as in Specimens S2, and consequently, promote link web fracture. However, Specimen S4 sustained one more loading cycle than Specimen S2 before failing by link web fracture. Therefore, two-sided fully welded stiffeners were no more detrimental to the occurrence of web fracture than one-sided stiffeners. Specimen S6 had the same stiffener arrangement as Specimen S4, but used a different welding process to weld the stiffeners to the link flanges and web. Specimen S4, which used the shielded metal arc welding (SMAW) process with an E7018 electrode, achieved greater rotation and survived four-and-one-half more loading cycles than Specimen S6, which used the self shielded flux cored arc welding (FCAW) process with an E70T-6 electrode. The lower fracture toughness and higher heat input of the FCAW process may have caused the degraded performance of Specimen S6 compared to Specimen S4. Specimen S5 was provided with unconventional link stiffener detailing. This specimen had full depth stiffeners on both sides of the web, but these stiffeners were welded only to the flanges and not to the web. Eliminating the stiffener to link web weld was expected to avoid the occurrence of fracture in the k-area of the link web. In the absence of stiffener to link web welds, it was expected that restraint against link web buckling would be achieved by “sandwiching” the link web between stiffeners on opposite sides of the web. Unlike specimens with fully welded stiffeners, Specimen S5 exhibited notable web buckling starting from early inelastic loading cycles. Although no fracture occurred near the k-area of the section, this specimen ultimately fractured at a location of concentrated buckling deformation, where the link web was rubbing against the stiffener. Specimen S5, which had stiffeners welded to the link flanges only, completed three more loading cycles and achieved a larger inelastic rotation than Specimen S2, which was provided with conventional stiffener detailing. This approach of sandwiching the link web between stiffeners without applying welds to the web can provide excellent cyclic performance, and merits further investigation. Specimen S7 was also provided with unconventional stiffener detailing. For this specimen, partial depth stiffeners were provided on only one side of the web, and were welded only to the web and not to the flanges. This arrangement

was expected to reduce the local restraint near the k-area, although it was questionable whether the stiffeners would provide sufficient buckling restraint for the web. Specimen S7 developed a smaller inelastic rotation than Specimen S2, and failed due to web fracture at the termination of stiffener welds. Buckling deformation of the web concentrated near the vertical ends of the stiffeners, and eventually caused fracture of the web. Therefore, eliminating the connections of the stiffeners to the flanges was not beneficial for mitigating fracture development in the web. Specimens S5 and S6 were duplicated and retested using the revised loading protocol. These two specimens, designated respectively as S8 and S9, developed inelastic rotations much greater than the required 0.08 rad. The failure modes were similar to the failure modes of Specimens S5 and S6. These results suggest that the two-sided stiffeners, whether welded or not welded to the link web, are effective arrangements to control link rotation capacity. Specimen S10 was provided with fewer stiffeners (i.e., larger stiffener spacing) than required in the 2005 AISC Seismic Provisions. Under the revised loading protocol, this specimen developed an inelastic rotation much greater than the required 0.08 rad, and failed due to fracture of the link web. Although the specimen exhibited strength degradation caused by substantial web buckling, Specimen S10 ultimately failed due to link web fracture. This test result suggests that the shear link stiffener spacing requirement in the Provisions is conservative. Further research may be beneficial to determine if a relaxation in the stiffener spacing criteria is justified. Based on the above observations for Specimen S10, the link web fractures which were observed in the current tests, but not reported from earlier tests, do not appear to be the result of the stiffener spacing criteria in the Provisions. Alternative stiffener designs which are not currently permitted by the Provisions, such as the sandwiching stiffeners used in Specimens S5 and S8, may be effective for link detailing. As demonstrated by the comparison between Specimens S4 and S6, the welding process used for the stiffener to web weld may have a significant influence on the link web fracture. 5.3. Material properties The proximity of the link web fractures in many of the test specimens to the k-line of the section suggests that material properties in the k-area may have played a role in these fractures. The commentary on the 2005 AISC Seismic Provisions discusses that the steel in the k-area of rolled wideflange shapes (the region where the web meets the flange) can exhibit high hardness and be prone to fracture. In order to investigate the effect of material properties on link behavior, specimens of identical geometry, Specimens S1, S2, and S3, were constructed from W10×33 shapes produced by three different mills. The three steels are indicated in Table 1 as (A), (B), and (C). Sample ASTM Rockwell B hardness measures taken along the web-centerlines for the three steel sections are shown in Fig. 10(a). While all three sections show notably higher hardness values near the k-area of the section,

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performance between the three specimens, S1, S2, and S3. Due to the higher hardness and lower elongation in their k-areas, Specimen S1 (Steel (A)) and S3 (Steel (C)) were expected to exhibit a smaller rotation capacity compared to Specimen S2 (Steel (B)). However, the overall performance of the three specimens was quite similar, with Specimen S3 completing three more loading cycles than Specimens S1 and S2. All three specimens failed by fracture of the link web initiating at the stiffener weld terminations. While these test results indicate that the steel material can influence link performance, the results do not clarify the correlation between the reduced material ductility in the k-area and the occurrence of link web fracture. It is possible that the difference in k-area material properties between the three steels, which were all A992 steel, was insufficient to clearly highlight the influence of k-area material properties. A rather large number of tests in this program indicate that the loading history and stiffener details do not change the failure mode of shear links from that controlled by link web fracture to a different failure mode. However, the relation between the material properties and the link web fracture is less clear. It is suggested that further studies be conducted to clarify the effect of material properties, specifically the properties in the k-area, on link web fracture. 6. Link end welds

Fig. 10. Material properties measured for W10×33 sections: (a) hardness measures along web centerline of link section; and (b) tension coupon test results for Steel (C).

Steel (A) shows the highest peak hardness value, while Steel (B) shows somewhat lower hardness values. The peak value is seen at a distance of 20–25 mm from the outer face of the flange, at the k-line of the section. Tension coupons were taken from the location of the web indicated in Fig. 10(a). For Steels (A) and (C), the coupons taken from the k-area showed 25–35% higher tensile strength and a two-thirds reduction in elongation compared to the coupon taken from the mid-depth of the web. For Steel (B), the coupons taken from the k-area showed a 5% higher tensile strength and one-third reduction in elongation compared to the coupon taken from the mid-depth of the web. Fig. 10(b) shows the tensile coupon test results for Steel (C), comparing three coupons taken from the edge of the flange, mid-depth of the web, and k-area of the section. The figure illustrates the significantly higher tensile strength and reduced elongation of the coupon taken from the k-area. Elevated hardness values, elevated tensile strength, and reduced ductility is characteristic of the k-area of roller straightened shapes [19], and was observed in all sections used for the link specimens in this study with the exception of the W10×19. For specimens in Group III, the stiffener weld was terminated within the region of elevated hardness values, as indicated in Fig. 10(a), so that the degraded material properties would influence their performance. The notable difference in the k-area material properties was expected to cause varying

The link specimens tested in this program were welded to heavy end plates, which in turn were bolted into the test setup shown in Fig. 1. Studying the behavior of connections between the link and surrounding members was not an objective of this research program. Consequently, the end plate connection details used in these tests were not intended to represent realistic link end conditions. Rather, the end plate detail was devised to preclude failure at the link ends, and thereby to permit study of link behavior. Nonetheless, the results of this test program provide some useful insights into potentially effective link-to-column connection details. As discussed earlier, Specimens 1A, 1B, and 6A failed prematurely due to fracture of a fillet weld connecting the link flange to the end plate. The fracture of the end plate welds occurred either in the throat of the weld, or in the link flange base metal near the weld–base metal interface. Similar fractures were reported in tests by Kasai and Popov [7] and Ramadan and Ghobarah [20]. In order to avoid these fractures, in later specimens, the leg size of the fillet welds was increased to oneand-one-half times the link flange or web thickness (see Fig. 2). Further, weld tabs were used at the edges of the flanges, as shown in Fig. 11, to avoid introducing undercuts or weld defects at these edges. With the exception of Specimen 1A, which used the FCAW process with an E70T-6 electrode, the fillet welds were made using the SMAW process with an E7018 electrode. Instead of using fillet welds, the 20 mm-thick flanges of the W10×68 links were connected to the end plates with partial penetration groove welds (groove depth of 14 mm), reinforced at the root side of the weld by fillet welds with a leg size equal to the flange thickness.

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Fig. 11. Typical end plate weld.

The improved end plate welds shown in Fig. 11 were used at both ends of thirty-one link specimens. No damage in the end plate welds was observed in any of these specimens. Therefore, although the thickness of the end plates of 50 mm (three to five times the thickness of the link flange) may be unrealistic for practical application, the end plate connections used in these tests are promising for application to link-tocolumn connections. Research is continuing by the authors to develop practical link-to-column connection details. 7. Design implications 7.1. Link flange slenderness limit A basic objective of this research program was to study the effect of flange slenderness on link behavior. More specifically, the objective was to determine if the flange slenderness limit, b f /2t f , for link flanges can be increased from the seismically compact limit of 0.30(E/Fy )1/2 to the compact limit of 0.38(E/Fy )1/2 . In this test program, the flange slenderness of the W10×33 was very close to 0.38(E/Fy )1/2 for Fy = 345 MPa. Specimens with W10×33 sections were tested over a range of lengths varying from 1.0 to 3.6M p /V p , covering a wide range of link behaviors ranging from shear to flexural dominated response. The longer W10×33 specimens (Specimens 6B and 7) performed very well. These specimens exhibited flange buckling, but significantly exceeded their required rotation levels before link strength dropped below the defined failure threshold. The shorter W10×33 specimens tested using the old-AISC loading protocol (Specimens 4A to 4C and 5) did not achieve their required rotations due to premature web fractures. However, when retested using the more realistic revised loading protocol, short W10×33 specimens (Specimens 4A-RLP and 4C-RLP) significantly exceeded their required rotation. Similar results were obtained from Specimens S1 through S10. Therefore, data from the W10×33 test specimens suggest that the compact limit for flange slenderness is adequate for links of all practical lengths. Companion finite element simulations of EBF links by Richards

and Uang [15], calibrated to the results of these tests, and extended to a wider range of link parameters, also support the use of the compact limit for all link lengths. The specimens constructed with W16×36 sections also provide useful data on flange slenderness effects. Based on nominal section dimensions, the flange slenderness of the W16×36 falls between the seismically compact limit and the compact limit. However, based on measured dimensions, the actual flange slenderness was smaller than the nominal value, and fell just within the seismically compact limit. The failure mode of Specimens 8 and 8-RLP, which were W16×36 links with a length of e = 1.5M p /V p , was unique compared with other shear yielding links. The specimens developed significant flange and web buckling at both ends of the link, and ultimately failed by web fracture at the link end panels. Although the flange and web buckling did not directly cause strength degradation, it appeared that the severe web buckling triggered rapid growth of the web crack. Specimens 9 and 9-RLP, which were W16×36 links with a length of e = 2.0M p /V p , failed due to strength degradation associated with combined flange and web buckling in the link end panels. Between the two W16×36 links tested using the revised loading protocol, Specimen 8-RLP significantly exceeded the required rotation, while Specimen 9-RLP only barely met the rotation requirement, as discussed earlier. The tendency of local buckling in the W16×36 sections suggests caution in permitting the compact limit of 0.38(E/Fy )1/2 for the flanges of longer flexure dominated links. The strength degradation in the W16×36 links appeared to result from flange–web interaction, and further studies of such interaction are needed. There is strong and consistent evidence from the results of this testing program, results from analytical studies [15], as well as results from previous tests [7], to support the less stringent limit of 0.30(E/Fy )1/2 for shear yielding links. For longer links (e > 1.6M p /V p ), the evidence on flange slenderness effects on link rotation capacity is not as clear. A number of longer link specimens with a flange slenderness at the limit of 0.38(E/Fy )1/2 provided excellent performance, achieving inelastic rotations well beyond the required levels. However, rather short specimens (e = 1.5 and 2.0M p /V p ) constructed with the W16×36 section showed a notable tendency for flange buckling. Since the flange slenderness of the W16×36 was within the limit of 0.30(E/Fy )1/2 , it is unclear if the flange slenderness limit for links dominated by flexure could be relaxed to 0.38(E/Fy )1/2 . Based on this research and others, the flange slenderness limit for shear links (e ≤ 1.6M p /V p ) has been relaxed to 0.38(E/Fy )1/2 in the 2005 AISC Seismic Provisions. 7.2. Inelastic rotation limit The loading protocol used for testing has a very significant effect on the performance of link specimens. This can be seen in Fig. 12(a), which compares the inelastic rotation measured for all twenty-four shear yielding links tested in this program. Although the data is skewed by the large number of W10×33 links (fifteen specimens) and W18×40 links (five specimens),

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Fig. 12. Performance of shear yielding links: (a) inelastic rotation; and (b) overstrength factor.

each of identical length, the figure indicates the inelastic rotation is significantly affected by the loading protocol used for testing. The average inelastic rotation was 0.075 rad for the six specimens tested with the old-AISC protocol, 0.117 rad for the eight specimens tested with the revised protocol, and 0.062 rad for the eight specimens tested with the severe protocol. While the influence of link section geometry, material, and stiffener details may also be recognized from the figure, the influence of loading protocol is dominant over these secondary factors. The data in Fig. 12(a) emphasize the importance of choosing a loading protocol that realistically reflects expected earthquake demands, such as the revised loading protocol for EBF links developed by Richards and Uang [16,17]. This revised loading protocol has been adopted by the 2005 AISC Seismic Provisions [1]. All shear yielding links tested in this program with the revised loading protocol consistently exceeded the required inelastic rotation of γ p = 0.08 rad. With the exception of Specimen 9, the intermediate and flexural yielding links, tested with the old-AISC protocol, met their inelastic rotation requirements. Specimen 9-RLP, which was a duplicate of Specimen 9, tested with the revised protocol, exceeded its required rotation. Analysis by Richards and Uang [17] suggests that while the old-AISC protocol was overly conservative for

shear yielding links, it was adequate for flexure yielding links. Therefore, data from the current project suggest that links designed according to the 2005 AISC Seismic Provisions would reliably develop the prescribed inelastic rotation. 7.3. Link overstrength The link overstrength values evaluated for each specimen are presented in Table 4. This table lists the ratio Vmax /Vn , where Vmax is the largest shear force measured in a test. Vn is the plastic strength of the link, and was calculated per the 2005 AISC Seismic Provisions as the smaller of V p or 2M p /e, where V p and M p were computed using the actual measured dimensions and actual measured yield strengths of the test sections. As indicated in Figs. 4, 6 and 8, the measured value of Vn was typically equal to or somewhat larger than its nominal value (based on nominal dimensions and nominal yield strength). The overstrength tended to be greater for shorter links of lengths between 1.0 and 1.7M p /V p , compared to longer links with a length greater than 1.7M p /V p . The average overstrength for these shorter link specimens was 1.41, with a variation from 1.25 to 1.62. The average overstrength for the longer specimens (e > 1.7M p /V p ) was 1.20, with a variation from 1.05 to 1.27.

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Table 4 Specimen overstrength Specimen

Vmax /V p

Specimen

Vmax /V p

1A 1B 1C 2 3 4A 4B 4C 5 6A 6B 7 8 9 10 11 12

1.20 1.20 1.23 1.24 1.26 1.40 1.42 1.41 1.34 1.21 1.21 1.27 1.35 1.11 1.44 1.42 1.40

4A-RLP 4C-RLP 8-RLP 9-RLP 10-RLP 11-RLP 12-RLP 12-MON 12-SEV 12-RAN S1 S2 S3 S4 S5 S6 S7 S8 S9 S10

1.45 1.47 1.37 1.05 1.47 1.42 1.44 1.59 1.36 1.62 1.49 1.29 1.56 1.37 1.25 1.32 1.28 1.27 1.43 1.36

Fig. 12(b) compares the overstrength factor measured for all shear yielding links tested in this program. The average link overstrength was 1.41 for the specimens tested with the old-AISC protocol, 1.41 for the specimens tested with the revised protocol, and 1.37 for the specimens tested with the severe protocol. Specimens 12 and 12-RLP, 12-SEV, 12-MON, and 12-RAN were subjected to various different loading histories. Specimen 12-SEV, which was subjected to a large number of inelastic loading cycles before failing at a rotation of γ p = 0.072 rad, developed an overstrength of 1.36, whereas the monotonically loaded Specimen 12-MON achieved an overstrength of 1.59 at a rotation of roughly γ p = 0.21 rad. Comparing the seven identical specimens (five of which are shear yielding links) tested with both the old-AISC and revised loading protocol, the overstrength was 1.36 for the specimens tested with the old-AISC protocol, and 1.38 for the specimens tested with the revised protocol. The data suggest that the loading protocol has a more limited influence on link overstrength than on link rotation. Specimens 10, 11, 10-RLP, and 11-RLP were made of the W10×68 section, which had a high ratio of flange to web area. These specimens were expected to develop shear resistance in the flanges, and therefore, greater overstrength compared to the other specimens. However, these specimens did not show unusually large values of link overstrength compared to other shear link specimens. A notably large variation in overstrength values is seen between Specimens S1, S2, and S3. These three specimens were W10×33 links of identical length, tested with the severe loading protocol, but constructed from different steels (see Table 1 and Fig. 10). The higher overstrength values of Specimens S1 and S3 compared to Specimen S2 may be partly due to the significantly higher tensile strength measured in the k-area of the section, which was not accounted for in evaluating the nominal shear strength. However, since the

k-area would represent a smaller fraction of the web depth, the higher strength in the k-area should have a more limited influence on the overstrength factor of deeper sections. The data for Specimens S2 and S4 through S7, which were constructed from the same steel but provided with different stiffener details, suggest that the stiffener details have little influence on the overstrength factor. While the test data suggest smaller overstrength factors for longer links in the length range of e > 1.7M p /V p , the W16×36 links showed especially low overstrength values. The overstrength values for Specimens 9 and 9-RLP, both with a length of e = 2.0M p /V p , were 1.11 and 1.05, respectively. Combined flange and web buckling caused drastic strength degradation in these specimens. In fact, links with a length near e = 2.0M p /V p are expected to experience significant shear-flexure interaction [14], and thus, the inelastic strength as defined by the 2005 AISC Seismic Provisions is likely to be conservative for these links. The occurrence of local buckling and overestimation of the inelastic strength were the likely reasons for the particularly low overstrength values for Specimens 9 and 9-RLP. 8. Conclusions This paper summarized an experimental program on the cyclic loading behavior of EBF links made of ASTM A992 steel. The test results provide data pertaining to a wide range of design issues for EBFs including the flange slenderness limits, overstrength factors, and stiffener design for links. Results of this test program clearly show that the loading protocol used to test EBF links has a very large effect on the inelastic rotation achieved by the links. Since the loading protocol has such a large effect on link test results, it is important that loading protocols realistically reflect demands caused by actual earthquake loading. Such a loading protocol for EBF links was recently developed by Richards and Uang [16,17], and has been adopted by the 2005 AISC Seismic Provisions. A number of shear yielding links tested in this program failed due to fracture of the link web. These fractures initiated at terminations of stiffener to link web fillet welds and ultimately caused rapid strength degradation. This type of link web fracture has not typically been observed in earlier link tests reported in the literature. Observations from this program suggest that the loading history and stiffener arrangement have limited influence on the link web fracture. Further investigation is recommended to study the effect of material properties in the k-area on the occurrence of the link web fracture. Link web fracture can be delayed and link rotation capacity can be enhanced by altering the stiffener details. One method is to increase the distance from the k-line of the rolled link section to the termination of the stiffener to link web fillet weld. Based on the test results, it is recommended that stiffener welds be terminated a distance of at least five times the web thickness from the k-line of the link section. Another method to delay web fracture is to restrain both sides of the link web using stiffeners without placing welds directly to the web. The stiffeners are

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welded only to the flanges and not to the web. This technique, while promising, requires further investigation. Test data from this program indicate that the flange slenderness limit of shear yielding links (e ≤ 1.6M p /V p ) can be relaxed to the compact limit of 0.38(E/Fy )1/2 . For longer links (e > 1.6M p /V p ), it is recommended that the flange slenderness limit be maintained at the seismically compact limit of 0.30(E/Fy )1/2 , pending further study of the effects of shear-flexure interaction on local buckling. All link specimens conforming to these flange slenderness limits were capable of achieving the inelastic rotations required in the 2005 AISC Seismic Provisions. The ASTM A992 rolled wide-flange links tested in this program exhibited overstrength factors ranging from 1.05 to 1.62, with an overall average of 1.35. Sections with high ratios of flange to web areas did not exhibit unusually high overstrength factors, at least within the range of flange to web area ratios typical of rolled wide-flange shapes. The overstrength factor of 1.5, which forms the basis for the capacity design procedure in the 2005 AISC Seismic Provisions, appears reasonable for links constructed of typical rolled shapes. However, based on experimental and analytical results reported by others, a higher overstrength factor may be appropriate for short links constructed of built-up shapes with heavy flanges [15]. As a closing remark, the large number of tests conducted in this research program suggests that EBF links constructed of ASTM A992 steel and designed according to the 2005 AISC Seismic Provisions perform well, and meet the performance requirements of the 2005 AISC Seismic Provisions. Acknowledgements The writers gratefully acknowledge primary funding provided for this project by the American Institute of Steel Construction (AISC) and the National Science Foundation (Grant No. CMS-0000031). The first author expresses gratitude for sponsorship provided by the Twenty-First Century Center of Excellence Program awarded to the Tokyo Institute of Technology, Japan. The tests discussed herein were conducted as Masters’ thesis work by former students at the University of Texas at Austin, Gabriela Arce, Han-Choul Ryu, and Pedro Galvez. The writers would like to particularly thank Tom Schlafly of AISC for his support and assistance throughout this project. The writers thank Chia-Ming Uang, Paul Richards, James Malley, Subhash Goel, and Tom Sabol for their assistance and advice on this study. References [1] American Institute of Steel Construction, Inc. (AISC) . Seismic provisions for structural steel buildings. Standard ANSI/AISC 341-05. Chicago (IL, USA): AISC; 2005.

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[2] Popov EP, Engelhardt MD. Seismic eccentrically braced frames. Journal of Constructional Steel Research 1988;10:321–54. [3] Itani AM, El-Fass S, Douglas BM. Behavior of built-up shear links under large cyclic displacement. Engineering Journal, American Institute of Steel Construction 2003;40(4):221–34. [4] McDaniel CC, Uang C-M, Seible F. Cyclic testing of built-up steel shear links for the new bay bridge. Journal of Structural Engineering, American Society of Civil Engineers 2003;129(6):801–9. [5] Hjelmstad KD, Popov EP. Seismic behavior of active beam link in eccentrically braced frames. Report No. UCB/EERC-83/15. Berkeley (Richmond, CA, USA): Earthquake Engineering Research Center, University of California; 1983. [6] Malley JO, Popov EP. Design considerations for shear links in eccentrically braced frames. Report No. UCB/EERC-83/24. Berkeley (Richmond, CA, USA): Earthquake Engineering Research Center, University of California; 1983. [7] Kasai K, Popov EP. A study of seismically resistant eccentrically braced frames. Report No. UCB/EERC-86/01. Berkeley (Richmond, CA, USA): Earthquake Engineering Research Center, University of California; 1986. [8] American Institute of Steel Construction, Inc. (AISC) . Seismic provisions for structural steel buildings. Standard ANSI/AISC 341-02. Chicago (IL, USA): AISC; 2002. [9] Okazaki T, Arce G, Ryu H-C, Engelhardt MD. Experimental study of local buckling, overstrength, and fracture of links in eccentrically braced frames. Journal of Structural Engineering, American Society of Civil Engineering 2005;131(10):1526–35. [10] American Institute of Steel Construction, Inc. (AISC) . Specification for structural steel buildings. Standard ANSI/AISC 360-05. Chicago (IL, USA): AISC; 2005. [11] Arce G. Impact of higher strength steels on local buckling and overstrength in eccentrically braced frames. Master’s thesis. Austin (TX, USA): Department of Civil Engineering, University of Texas at Austin; 2002. [12] Ryu H-C. Effects of loading history on the behavior of links in seismicresistant eccentrically braced frames. Master’s thesis. Austin (TX, USA): Department of Civil Engineering, University of Texas at Austin; 2005. [13] Galvez P. Investigation of factors affecting web fractures in shear links. Master’s thesis. Austin (TX, USA): Department of Civil Engineering, University of Texas at Austin; 2004. [14] Engelhardt MD, Popov EP. Behavior of long links in eccentrically braced frames. Report No. UCB/EERC-89/01. Berkeley (Richmond, CA, USA): Earthquake Engineering Research Center, University of California; 1989. [15] Richards P, Uang C-M. Evaluation of rotation capacity and overstrength of links in eccentrically braced frames (phase 1). Report No. SSRP2002/18. La Jolla (CA, USA): Department of Structural Engineering, University of California at San Diego; 2002. [16] Richards P, Uang C-M. Development of testing protocol for short links in eccentrically braced frames. Report No. SSRP-2003/08. La Jolla (CA, USA): Department of Structural Engineering, University of California at San Diego; 2003. [17] Richards P. Cyclic stability and capacity design of steel eccentrically braced frames. Ph.D. dissertation. La Jolla (CA, USA): University of California, San Diego; 2004. [18] Krawinkler H, Gupta A, Medina R, Luco N. Loading histories for seismic performance testing of SMRF components and assemblies. Report No. SAC/BD-00/10. Sacramento (CA, USA): SAC Joint Venture; 2000. [19] Miller KR, Frank K. Study of the material properties of the web–flange intersection of rolled shapes. Report No. SAC/BD-99/08. Sacramento (CA, USA): SAC Join Venture; 1999. [20] Ramadan T, Ghobarah A. Behavior of bolted link–column joints in eccentrically braced frames. Canadian Journal of Civil Engineering 1995; 22:745–54.