Experimental study of steel-timber composite beam-to-column joints with extended end plates

Construction and Building Materials 226 (2019) 636–650

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Construction and Building Materials journal homepage: www.elsevier.com/locate/conbuildmat

Experimental study of steel-timber composite beam-to-column joints with extended end plates Abdolreza Ataei a,b,⇑, Hamid R. Valipour b, Mark A. Bradford b, Alireza A. Chiniforush c a

Department of Civil Engineering, The University of Isfahan, Iran Centre for Infrastructure Engineering and Safety, School of Civil and Environmental Engineering, UNSW Sydney, NSW 2052, Australia c School of Civil Engineering, The University of Sydney, NSW 2008, Australia b

h i g h l i g h t s  Presents deconstructable and sustainable steel–timber composite joint.  Structural behaviour of steel-timber composite joints were reported in details.  Joint has very favourable moment and rotation capacities.  Adding CLT panels to the bare steel joints has a significant effect on their behaviour.  Rigid plastic design is applicable for this joint.

a r t i c l e

i n f o

Article history: Received 13 October 2018 Received in revised form 13 May 2019 Accepted 17 July 2019

Keywords: Bolted shear connectors Deconstruction Semi-rigid beam-to-column joint Steel-timber composite joint Sustainable construction

a b s t r a c t This study presents the results of laboratory push-down tests conducted on one pure steel and four steeltimber composite (STC) cruciform subassemblies to assess the failure characteristics, stiffness, flexural resistance and ductility of the extended end plate STC beam-to-column connections subjected to negative (hogging) bending moment. In the proposed composite system, the cross -laminated timber (CLT) panels were connected to the top flange of steel girders using coach screws and the steel beams were connected to the steel columns by bolted extended end plates. Moreover, the two juxtaposed CLT slabs (subject to tension) were connected by the mechanically anchored threaded rod and/or surface spline joints with steel plates. The experimental results showed that the extended end plate STC connection have enough rotation capacity to provide for plastic analysis/design of the STC beams. Furthermore, it was shown that the composite action in conjunction with continuity of timber slab can increase the bending moment capacity of the connection more than 50% of that for a pure steel connection. Ó 2019 Elsevier Ltd. All rights reserved.

1. Introduction End Plate Semi-Rigid (EEPSR) steel and steel-concrete composite joints (Fig. 1) have been increasingly adopted in modern steel-framed buildings, especially in earthquake-prone regions, because of their relative ease of construction, superior structural performance and economical attributes compared to rigid connections. In addition, the extended end-plate connections have excellent rigidity, ductility and rotation capacity which are required for redistribution of bending moment during extreme loading scenarios. Accordingly, several laboratory experiments and numerical (finite element) simulations have been performed on end plate

⇑ Corresponding author. E-mail address: [email protected] (A. Ataei). https://doi.org/10.1016/j.conbuildmat.2019.07.154 0950-0618/Ó 2019 Elsevier Ltd. All rights reserved.

steel-concrete composite joints [1–6]. For example, Brown and Anderson [1] fabricated and tested five steel-concrete composite beam-to-column cruciform subassemblies with flush-end plate. The initial stiffness and hogging bending moment resistance of the connections were determined and it was concluded that even for cruciform subassemblies with relatively deep beams, the contribution of reinforced concrete slab in overall stiffness and bending moment capacity of the connection is significant [1]. To evaluate effect of the precast hollow-core concrete slabs on the structural performance of the semi-rigid beam-to-column connections, Fu and Lam [2] tested eight cruciform subassemblies. Effect of the slab thickness, amount of reinforcing bars in the slab and spacing of the shear studs/degree of shear connections were considered in the experimental program and a simple model for predicting bending moment and rotation capacity of the composite joints with hollow-core concrete slabs was proposed [2]. Gil et al.

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Fig. 1. Typical extended end plate joints (a) bare steel and (b) composite joint.

[3] fabricated two interior and an edge 3D beam-to-column joints to evaluate effect of the minor axis semi-rigid connections and loading on the structural performance of the steel-concrete composite beam-to-column connections in the major direction. Furthermore, detailed nonlinear 3D finite element (FE) models of the 3D subassemblies were built and effect of beam size on the structural performance of the connections were studied [3]. Most of the experimental studies [1-3] so far have focused on composite steel beams made of I-sections welded to an end plate that in turn was bolted to a column with H-section. In these studies, the premature failure of the flange and web of the columns (H-section) were prevented by using web stiffeners welded to the column web and flanges at the level of the bottom and top flanges of the beam. Moreover, behaviour of the composite I-beams with end plates bolted to the concrete-filled columns have been studied [4,5]. In the subassemblies with concrete filled steel tubes (CFST), the flush end plates are typically connected to the column by blind bolts [4,5]. In an experimental study by Loh et al. [4], five subassemblies with CFST columns were fabricated and tested to study effect of the level of shear interaction and amount of reinforcing bars on the structural performance of the beam-to-column connections. It was concluded that partial shear interaction in the hogging bending moment zones of the composite floors can increase the rotation capacity and ductility of the semi-rigid connections, with only a minor compromise of stiffness and bending moment capacity [4]. Mirza and Uy [5] tested three specimens, i.e. one bare steel and two steel-concrete composite subassemblies with CFST columns. The first two specimens were subjected to monotonically increasing static loads and the third specimen was subjected to cyclic loads. The results of test revealed acceptable bending moment, rotation capacity and seismic performance of the composite beam-to-CFST column connections with flush end plates [5]. Moreover, it was shown that detailed 3D finite models of the subassemblies can predict the cyclic load-displacement behaviour of the subassemblies with very good accuracy. In the past decade, significant emphasis has been placed on development of sustainable methods and technologies that potentially lower carbon emissions, demolition wastes and raw material usage by improving the deconstructability of the structures and accordingly facilitating the material recycling in the construction industry. However, the conventional steel-concrete composite joints are energy-intensive and environmentally-intrusive and life cycle assessments (LCA) have shown that replacing conventional reinforced concrete floors with more sustainable alternatives has the highest or second highest potential for savings in embodied energy and carbon [7]. Recently, a cradle to grave approach was adopted for the LCA of mid- to high-rise (5 to 15 storey) residential and office buildings with different structural systems (reinforced concrete, steel-concrete composite and steel-timber composite) and it was shown that replacing conventional concrete slabs with

cross-laminated timber (CLT) panels can nearly halve the embodied energy associated with the structure [8]. The conventional steel-concrete composite floors are also wasteful and not conducive to deconstruction and recycling/reusing of the construction materials [8–11]. Because, the composite action in conventional composite floors is typically provided by means of welded stud shear connectors permanently embedded within in-situ concrete slabs that hinders dismantling, recycling and reusing of the structural floors. This shortcoming of conventional steel-concrete composite floors has been addressed by bolted shear connectors and several promising studies have been conducted on sustainable steel-concrete composite joints with bolted shear connectors that can improve the deconstructability and reusability of conventional composite floor systems. Extensive experimental and numerical studies on composite cruciform beam-to-column joints and composite beams have been conducted at UNSW Sydney to investigate the possibility of using deconstructable post-installed bolted shear connectors. In these studies, cruciform beam-to-column subassemblies with (open) I-section columns [12,13] and concrete-filled steel tube columns [14,15] made of mild 300PLUS and 350 steel grade have been fabricated and tested under a monotonically increasing static load. Moreover, cruciform subassemblies with high strength steel (HSS) flush end plates [16,17] have been fabricated and subjected to monotonically increasing displacement-controlled load to determine the negative (hogging) bending moment capacity, rotation capacity and ductile/ brittle failure modes of the deconstructable composite beam-tocolumn connections. It is noteworthy that testing of cruciform subassemblies which consist of two beams framed into opposite sides (typically flanges) of a column have been widely used for evaluating the structural performance of the interior beam-to-column connections in steel and steel-concrete composite floors [18]. Apart from the beam-to-column joints, full-scale steel-concrete composite beams with post-installed bolt shear connectors have been fabricated and tested under 4-point bending to evaluate the structural behaviour of the deconstructable composite beams subjected to sagging bending moments [11,19]. In all the beams and subassemblies tested, the composite action between the steel beam and the concrete slabs was provided by deconstructable bolted shear connectors. Using bolt shear connectors and bolted beam-to-column joints in steel-concrete composite frames results in a structure that can be deconstructed, and which has favourable credentials from an environmental perspective. These frames do, however, have the disadvantage of being heavyweight compared to recently developed composite floors that take advantage of prefabricated timber slabs/panels. The weight per unit volume of timber (i.e. ctimber = 4–6 kN/m3) is significantly smaller than that of the concrete (cconcrete = 22–25 kN/m3) and replacing the conventional concrete slabs with prefabricated timber slabs significantly reduces

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the self-weight of the building that in turn lower the earthquake induced inertial forces and improve the performance of the building in the seismic prone zones [20]. The tensile strength of timber is significantly higher than that of the concrete and, from environmental point of view, timber (wood) sequestrates carbon dioxide from atmosphere [8,21]. To explore these advantages, CLT panels were connected to the top flange of the steel beams in a cruciform beam-to-column subassembly with the extended end plate and the subassemblies were subjected to a push-down load to assess structural behaviour of the STC connections as a sustainable alternative to steel-concrete composite joints. The steel-CLT composite shear connection and members (beams and joints) comprising of CLT panels connected to top flange of the I beams have been the subject of ongoing research over the past few years [22–28]. Laboratory experiments and numerical simulations of several innovative steel-CLT composite connections have been conducted by Loss et al. [22–24] to investigate the feasibility and performance of the STC system. Hassanieh et al. [25] studied the structural behaviour of the hybrid STC beams by performing bending tests on seven steel-timber composite beams In this study, laminated veneer lumber (LVL) panels were connected to steel beams using different types of mechanical shear connectors, viz. high-strength bolts, coach screws or combinations of screws and glue. Additionally, seven steel-timber composite beam tests comprising of CLT panels connected to the top flange of steel beams were undertaken and reported by Hassanieh et al. [26] to determine the failure mode and evaluate the stiffness, loading capacity and ductility index of the steel-CLT composite beams. In these sets of tests, the CLT panels were attached to the steel beams using various mechanical fasteners (e.g. bolts, screws etc) and/or glue. The FE model of the STC beams was also developed and nonlinear analysis was performed too [26]. It was found that the peak loading capacity and stiffness of a bare steel beam is significantly enhanced, due to the development of composite action between the steel beam and the timber slab/panel. It was also concluded that all composite beams had a composite efficiency greater than 70%. Hassanieh et al. [28,29] also conducted experimental and numerical studies of steel-CLT composite connections with pushout tests in order to investigate the load-slip behaviour of these types of shear connection. However, to the best of authors’ knowledge, no experimental or finite element studies have been reported on the structural performance of the extended end plate STC beamto-column joints subjected to negative (hogging) bending moment regime. This paper presents the results of laboratory experiments conducted on four geometrically identical cruciform STC beamto-steel column subassemblies with bolted extended end plate connections. In the tested subassemblies, the CLT slabs/panels were connected to the steel girders using the coach screw shear connectors (Fig. 2) and the steel beams were connected to the steel columns by bolted extended end plates (see Fig. 1). A bare steel

Fig. 2. Cross-sectional configuration of shear connectors.

control specimen without a CLT slab was also fabricated and tested to assess the contribution of the CLT slab panels to the overall structural performance (i.e. stiffness and loading capacity) of the connections. The results of experiments were used to characterise the moment-rotation behaviour and evaluate rotational stiffness, rotation and bending moment capacities and ductility of this novel deconstructable composite joints. Moreover, effect of the type of CLT-to-CLT connection (i.e. surface spline with bolted steel plate and/or mechanically anchored threaded rods) across the column on the structural performance of the STC beam-to-column connections were investigated.

2. Laboratory experiment 2.1. Details of subassemblies In total, four (i.e. CJ2-CJ5) cruciform subassemblies (with STC beams) were tested to simulate and evaluate structural response of an internal joint in a hybrid floor with CLT slabs and semirigid extended end plate connections. In addition, one control joint (SJ1) without a slab was also fabricated and tested to assess the effect of composite action and presence of the CLT panels on the loading capacity and stiffness of the STC subassemblies under hogging bending moments. The outline of geometry and details of the bare steel and STC beam-to-column specimens are provided in Figs. 3 and 4, respectively. Furthermore, details of the CLT-to-CLT connections (i.e. surface spline joint and/or mechanically anchored thread rods) and size and spacing of the coach screw shear connectors are summarised in Table 1. The Australian 310UB40.4 steel beam and a 250UC72.9 steel column sections were used in the fabrication of the specimen. An extended end plate with 10 mm thickness was welded to end of each 310UB40.4 beam and the end plate was connected to the flange of the 250UC72.9 column with 8 M24 grade 8.8 bolts. All bolts were tightened using an electric torque wrench to ensure that the minimum Australian Standard specified post-tension force in the friction-grip bolts has been achieved. The minimum posttensioning force in the bolts was confirmed by Squirter Direct Tension Indicating (SDTI) washers. The size of steel beams, columns, thickness of the CLT slabs were selected with respect to the preliminary design of a hypothetical 12-storey office building with imposed action (live load) of 3 kPa [30]. The building was regular in plan with three 6.0 m long spans in each direction [30]. The structural steel frame were designed according to strength and serviceability limit state design requirement of AS4100 [31] and the thickness of the CLT slabs was determined with respect to the fire performance (considering 70 mm thick sacrificial layer) requirements and acceptable vibration performance criteria of the BS EN 1995-1-1 [32] standard. Each STC specimen was comprised of two CLT slab panels (Fig. 4) connected to the top flange of the 310UB40.4 steel beams using coach screws which had a length of 100 mm and diameter of 16 mm. The CLT slabs were 120 mm thick with five lamellae (30-20-20-20-30 mm) made of Spruce wood. The screws were installed through holes predrilled in the flange of the 310UB40.4 profile and the soffit of the CLT panels. The CLT panels were not continuous and to provide for development of significant hogging bending moments, the two juxtaposed CLT panels were connected by surface spline joints with steel plates (CJ2 and CJ3) and/or mechanically anchored rods (CJ4 and CJ5) as shown in Fig. 5. All subassemblies were geometrically identical except for the type of connection between the two CLT panels across the steel column. For specimens CJ2 and CJ3 having a steel plate connection, 16 mm diameter holes were marked on the CLT panels and steel

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Fig. 3. Geometric outline of joint SJ1.

Fig. 4. Geometry and details of joints CJ2-CJ5.

Table 1 Details of geometry and configuration of the beam-to-column subassemblies. Specimen

Surface spline plate

Threaded rod (mm)

ds (mm)

tep (mm)

shear connector

Spacing (mm)

SJ1 CJ2 CJ3 CJ4 CJ5

N.A. 600  250  6 600  250  6* – –

N.A. – – 16 12

N.A. 120 120 120 120

10 10 10 10 10

N.A. 16 mm screw

N.A. 200

Notes: ds = CLT panel thickness; tep = end plate thickness.

plates, and then drilled. The steel plates were then placed on the panels and high strength M16 8.8 bolts were installed and tightened (Fig. 5a & b). A 60  60  6 mm steel washer was also used (Fig. 5f) to spread the post-tensioning in the bolts and prevent crushing of the CLT panel around the bolt holes. For specimens CJ4 and CJ5, prior to mounting the CLT panels on the steel beams, holes and 70  70  70 mm pockets were marked and drilled in the panels. All threaded rods were then placed inside

the holes and tensioned (Fig. 5c & d). In the last step, the 70  70  70 mm pockets were filled with high-strength and low-shrinkage cementitious grout (Fig. 5e). To prevent crushing of the CLT in the anchorage zone, two 60  60  6 mm steel washers were used in conjunction with the threaded rods. The details and plan view of the CLT-to-CLT connections across the column are shown in Figs. 5 and 6. All nuts and bolts used in the CLT-toCLT connections were tightened to a snug-tight condition.

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Fig. 5. Types of connection between two CLT slabs (a) surface steel plate spline joint for CJ2; (b) surface steel plate spline joint for CJ3; (c, d & e) joints with mechanically anchored high-strength steel rods for CJ4 and CJ5 and (f) steel washer to prevent crushing of CLT.

2.2. Material tests All materials used in the experiments were ordered from the same batch. Details of the mechanical properties of the CLT, steel column and beam components, coach screws and extended end plates are provided in the following sections. 2.2.1. Steel profiles and plates The mechanical properties of the steel profiles and plates were determined following Australian Standard AS1391 [33] procedure. Direct tension tests were performed on the dog-bone coupons (three samples for each test) taken from the web and flange of the 310UB40.4 and 250UC72.9 profiles (300PLUS steel grade) and the steel plates to determine the elastic modulus and yield and ultimate strength of the steel profiles and plates. To precisely obtain the elastic modulus, extensometers were mounted on the dog-bone coupons within the elastic range during the tests. The

elongation of coupons over the gauge length was divided by the gauge length (of the extensometers) to determine the strain and plot the stress-strain diagrams of the steel coupons and accordingly calculate the elastic modulus and yield and ultimate strength of the steel. The strain rate for the uniaxial tension test kept at 0.0003 s 1 which was within the range specified by AS1391 [33]. Each uniaxial tension test took 7–10 min which was consistent with the time taken for testing each subassembly. The mean values of the mechanical properties and the stress-strain curve of the steel profiles and plates obtained from the direct tension tests are provided in Table 2 and Fig. 7a, respectively. 2.2.2. Cross laminated timber (CLT) The CLT slabs were 120 mm thick with five lamellae (30-20-2020-30 mm) made of C24 Spruce wood. Moisture content of the CLT panels was determined immediately after each test using the ovendry method of AS/NZS2098.1 [34]. The mean moisture content of

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Fig. 6. Outline of the CLT-to-CLT connections in (a) CJ2, (b) CJ3, (c) CJ4 and CJ5.

the CLT measured at the time of testing (by performing over-dry tests [35] on four samples taken from the CLT slabs of the STC subassemblies) was 13.0% (CoV = 0.1). The mean dry density of the CLT panels (for five samples) was 445 kg/m3 (CoV = 0.03) obtained

from AS/NZS 2098.7 [36] method. Four point bending tests were conducted on five identical CLT prisms according to CEN/EN 408 [37]. The mean bending strength of the CLT panels was fb = 33 MPa (CoV = 0.11) and the mean elastic modulus was E = 11.3 GPa

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2.3. Test setup and loading protocol

Table 2 Steel material tensile test results.

#

Specimen

Yield stress (MPa)#

Ultimate stress (MPa)#

Elastic modulus (GPa)#

Steel column flange Steel column web flange Steel beam flange Steel beam web Steel plate

329 [0.012]

525 [0.005]

192 [0.034]

352 [0.007]

523 [0.010]

202 [0.019]

352 [0.016] 399 [0.030] 325 [0.015]

531 [0.011] 551 [0.009] 495 [0.013]

200 [0.007] 219 [0.026] 203 [0.031]

Details of the test set up is shown in Fig. 8. All subassemblies were flipped over and placed upside down in the loading frame. The subassembly was placed on two roller supports 2.674 m apart (Fig. 8). To prevent local crushing of the CLT panels at the supports, a 10 mm thick 100 mm wide steel plate was placed between the rollers (at the supports) and the CLT slabs (Fig. 8). The pushdown displacement-controlled load was applied on a 10 mm thick steel plate welded to the top of the column (Fig. 8). The actuator (loading ram) attached to the steel plate (on top of the column) provided minimal rotational restraint for the column/specimens, and the end roller supports provided rotational freedom for the subassemblies during the test. All the tested subassemblies were symmetric with no sign of out-of-plane and/or accidental rotation (as evident from the inclinometers data) during the tests, thereby each side of the specimen was treated as a cantilever for the calculation purposes. The specimens were loaded monotonically using a hydraulic jack with 1000 kN capacity and maximum stroke of ±150 mm. The test setup, loading system and instrumentations at the commencement of each test was verified by applying a load corresponding to 10% of the estimated peak load carrying capacity and then unloading to zero. The loading protocol for the beam-tocolumn tests was according to BS EN 26891:1991 [39], which entailed each specimen being loaded initially to 40% of its estimated peak load over 120 s and the load being sustained at 40% of the estimated peak load (Fest) for 30 s (see Fig. 9). The specimen was then unloaded to 10% of its expected ultimate load and held constant for 30 s. In the last stage, the load was increased monotonically until failure (Fig. 9). The peak load carrying capacity of the specimens was roughly estimated using the results of previous tests conducted on beam-to-column connections with flush end plates [40,41]. Displacement rates of 0.3, 0.6 and 1.2 mm/min were used at different stages of loading. The test was concluded and loading was stopped, as soon as the vertical displacement of the column reached 150 mm (i.e. maximum stroke of the actuator) or the subassembly failed and the load (measured by the loadcell) dropped significantly (over 25% of the peak load).

The values in [] are the CoV of the test results (for three idential specimens).

(CoV = 0.09). The mean direct tensile strength of the CLT panels was ft = 25 MPa (CoV = 0.14) that was obtained from direct uniaxial tension tests on five 700 mm long dog-bone samples (240 mm gauge length, 250 mm shoulder width and 150 mm shoulder length). The bending and uniaxial tension tests were performed on CLT panels with average MC = 13%.

2.2.3. High strength steel bolts Tensile tests on the bolts were undertaken according to the Australian Standard AS4291.1 [38], with the cross-section of the bolts being reduced to the dimension specified in the standard and then tested under uniaxial tension loading. To determine mechanical properties (i.e. yield strength fy and ultimate strength fu) of the bolts, the uniaxial tension tests (with a strain rate within the range specified by AS1391 [33]) were conducted on three tensile coupon tests. The stress–strain plots of the 8.8 M24 bolts obtained from the uniaxial tension tests are provided in Fig. 7b. The mean yield and ultimate strengths of the grade 8.8 bolts were fy = 742 MPa (CoV = 0.016) and fu = 974 MPa (CoV = 0.018), respectively.

2.2.4. Coach screws Uniaxial tensile tests were conducted by Hassanieh et al. [26] according to AS4291.1 [38] to determine the yield stress, tensile strength and the stress–strain relationship for these shear connectors. The coach screws used in this study were made of Grade 4.6 steel, nominally having an ultimate tensile strength of 400 MPa and a yield strength of 240 MPa. The mean yield and ultimate strength of the coach screws (from three identical samples tested by Hassanieh et al. [26]) were fy = 305 MPa (CoV = 0.044) and fu = 430 MPa (CoV = 0.056), respectively. It is noteworthy that the coach screws used in Hassanieh’s experiments and in this study were from the same manufacturer.

2.4. Instrumentation To evaluate the structural performance and behaviour of the STC subassemblies with extended end plate connections, displacement/deflections, rotations and strains were respectively measured using Linear Strain Conversion Transducers (LSCTs), inclinometers and strain gauges mounted on the subassemblies (see Figs. 10 & 11).

600

1,000 400 300 Steel beam flange Steel beam web Steel column flange Steel column web

200 100

Stress (Mpa)

Stress (MPa)

500

800 600 Sample 1

400

Sample 2 200

Sample 3

0

0 0

50000 100000 150000 200000 Strain (μ )

0

50000 100000 Strain (με)

Fig. 7. Stress–strain plots of (a) steel profiles (b) flange and (b) grade 8.8 M24 bolts.

150000

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643

Fig. 8. Schematic test set up for the joint tests.

1.2 1

F/Fest

0.8 0.6 0.4

0.2 0 0

5

10

15

Time (min) Fig. 9. The loading protocol for the beam-to-column tests.

The vertical displacement at the centre of the column, separation of the two juxtaposed CLT panels in the specimens, the deformations of the extended end plates and the relative displacement (slip) at the interface between the CLT panels and the beams were measured by using LSCTs located at different locations (Fig. 10). Three inclinometers (with 0.01° precision) were also mounted on the specimens to measure the rotation at the different locations within the subassemblies. Two inclinometers (i.e. inclinometer-1 and 2 in Fig. 10) were attached to the web of the 310UB404.4 beam at the locations 50 mm away from the extended end plate and the third inclinometer was attached to the column panel zone (Fig. 10). The inclinometers were used to monitor the in-plane and any possible accidental out-of-plane rotations of the subassemblies. Furthermore, the difference between the rotations measured by inclinometer-1 and 2 was used to determine whether the deformations and rotations of the subassemblies have been

symmetric. The vertical displacement of the middle column was measured by an LSCT (i.e. LSCT-1) placed under the column (see Fig. 10). In order to measure differential movements (relative slips) between the CLT slab and the steel beam, the displacement transducers (i.e. LSCT-2 and 3 in Fig. 10) were attached to the steel beam by a magnet while the needle of LSCT was touching a steel plate fixed to the edge of the CLT panel. Moreover, LSCT-4/LSCT-5 on each side of the subassembly (see Fig. 11b) were used to measure the gap-opening between the two CLT slabs. The strains were measured at 16 locations along the steel beams and over the steel profile cross-section (notated as B1-B16 in Fig. 12a) and at 12 locations within the timber slabs (i.e. T1T12 in Fig. 11a and 12a). Moreover, seven strain gauges (EEP1EEP7) were installed on the extended end plates (Fig. 12b), four strain gauges (R1-R4) were mounted along the high-strength threaded rods in specimens CJ4 and CJ5 and two strain gauges (i.e. SP8-SP9 in Fig. 11a) were installed on the surface steel plates in the CLT-to-CLT slab connection zone of specimens CJ2 and CJ3. The steel gauges were 6 mm long and wood strain gauges were 50 mm long. The strain gauges were mainly placed on the regions closer to the column faces (viz. Fig. 11 and 12a) where higher strains/stresses were expected. The neutral axis depth and the strain distribution over the cross-section of the composite beam, and the attainment of the yielding strain can be established using the data from the gauges.

3. Test results and discussion 3.1. General In this section, the moment-rotation curves, the load-strain response along the beam axis, the load-strain response in the

Fig. 10. Layout of LSCTs and inclinometers.

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Fig. 11. LSCTs and strain gauges mounted on the (a) top surface of the CLT slab, (b) beam-to-column and CLT-to-CLT slab connection zone and (c) extended end plate.

Fig. 12. Layout of the strain gauges mounted on (a) steel beam and CLT slab and (b) extended end plates.

connection plates and rods, the load-strain response in the extended end plates and the load-end slip relationship at the middle and at the free ends of the beams are provided and discussed. It is noteworthy that the test results showed no vertical slip at the interface between the extended end plate and the column flanges, but, the extended end plate sustained significant (plastic) deformation at the region adjacent to the tensile flange. Moreover, the maximum difference between inclinometers-1 and -2 (Fig. 10) was less than 0.05° demonstrating the near symmetric deformation of the subassemblies.

3.2. Moment-rotation response The significantly nonlinear behaviour and relatively ductile behaviour (with credible rotation capacity) of the bare steel SJ1 subassembly and the four STC beam-to-column cruciform specimens are evident from the moment-rotation diagrams shown in Fig. 13. The maximum rotation capacity was considered as the rotation corresponding to the max. flexural capacity as per Eurocode 3 [32]. The bending moment for the connections was calculated by multiplying the reaction at the end support (i.e. one-half

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350

Moment (kN.m)

300 250 200

SJ1 CJ2 CJ3 CJ4 CJ5

150 100 50 0

0

20

40 60 Rotation (mRad)

80

Fig. 13. Bending moment vs rotation plots for the tested STC beam-to-column subassemblies.

of the applied load) and the distance between the support and the column face (i.e. 1.2 m as shown in Fig. 11). The rotation of connections was calculated by subtracting the rotation of the column (inclinometer 3) from the rotation of the beam (measured by inclinometers 1 and 2). The initial rotational stiffness ki, peak (maximum) bending moment capacity Mpeak and the corresponding rotational capacity hPeak required for characterising the behaviour of connections are summarised in Table 3. The final (failure) bending moment Mf and the corresponding rotation hf for all connections are also provided Table 3. In terms of the initial rotational stiffness ki, it is seen that the STC beam-to-column connections exhibited much higher rotation stiffness compared to the specimen SJ1 without any timber slab. The results in Table 3 demonstrate that the initial rotational stiffnesses of CJ2, CJ3, CJ4 and CJ5 were about 2.2, 2.1, 2.3 and 1.6 times that of SJ1, respectively. Therefore, it can be concluded that adding CLT panels to bare steel joints has a significant impact on the stiffness of the connection that should be considered under service loading conditions. Comparisons between specimens CJ3 and CJ2 or CJ5 and CJ4 also shows that the initial stiffness of the composite joints is significantly influenced by the stiffness of the CLT-to-CLT connection (i.e. size of the threaded rods and/or width of the surface steel plate). The rotational stiffness of the specimens significantly decreases due to the slip between the surface steel plates and CLT panels (in CJ2 and CJ3) and due to deformation of the CLT in conjunction with elongation of the threaded rods (in CJ4 and CJ5). The experimental results in Fig. 13 and Table 3 revealed that the peak bending moment capacity Mpeak of the STC subassemblies are much higher than that of the bare steel subassembly. The bending moment capacity of CJ2, CJ3, CJ4 and CJ5 was 1.45, 1.41, 1.20 and 1.31 times that of specimen SJ1, respectively. Hence, the presence of the timber slab improved the peak bending moment capacity of the extended end plate STC beam-to-column joints by at least 20%. The higher bending moment capacity of specimens CJ2 and CJ3

Fig. 14. Crushing (due to possible over tightening of the nuts and excessive posttensioning of the threaded rod) in conjunction with plug shear in CLT panels of specimen CJ4.

(compared to other specimens) should be attributed to higher strength of the CLT-to-CLT slab connections in CJ2 and CJ3. In terms of the rotation capacity and ductility, except for specimen CJ4, all joints (i.e. CJ2 to CJ5) have credible rotation capacities of 44.9, 48.2, 47.2, 22.2 and 41.5 mrad and the joints buckled locally after substantial rotation. The lower rotational and bending moment capacity of CJ4 is attributed to the failure mode of CJ4 which was associated with premature plug shear and crushing of the CLT panels in the anchorage zone (Fig. 14). According to EC3 [42] and EC4 [43] provisions, the rotation capacity of a connection must be bigger than 30 mrad for plastic analysis to be used. The test results showed that the CLT slabs can enhance the rotational capacity of the semi-rigid extended end plate connections above the minimum requirements set by the standards for plastic analysis of the steel frames. The stiffness of the connections with extended end plate was 300% and the bending moment capacity of the extend end plate connections was around 20% higher than that of the STC beam-to-column connections with flush-end plates [40,41]. 3.3. Strain in connection plates and rods The tensile strains in the surface CLT-to-CLT spline steel plates (Fig. 11a) are shown in Fig. 15. It is seen that the surface steel plate is subjected to large tensile stresses and yielding of plate occurs at around 85% of the peak load of the subassemblies. The tensile strains in the threaded rods (used for connecting the two CLT panels in specimens CJ4 and CJ5) were also measured and the loadstrain results are plotted in Fig. 16. Development of tensile strains/stresses in the connecting rods is evident from Fig. 16. Yielding of the rods in CJ5 occurred at around 75% of the ultimate

Table 3 Summary of the experimental results for the beam-to-column connections.

#

Specimen

Mpeak (kNm)

hPeak(mRad)

Mf (kNm)

hf (mRad)

ki (kNm/mrad)

smax# (mm)

Failure mode

SJ1 CJ2 CJ3 CJ4 CJ5

210.8 305.4 297.8 250.3 276.0

44.9 48.2 47.2 22.2 41.5

196.5 277.9 275.0 220.5 253.3

74.2 70.6 69.0 77.0 43.4

55.0 121.0 115.0 125.0 85.0

N.A. 4.8 3.8 2.8 1.7

Local buckling Local buckling Local buckling CLT fracture Local buckling

smax is the maximum slip between the CLT slab and the steel beam.

A. Ataei et al. / Construction and Building Materials 226 (2019) 636–650

Load (KN)

646

300

3.4. Strains in steel girders

250

The strains in the flanges of the steel girder recorded at two cross sections (i.e. 100 and 400 mm away from the face of the end plate) for all subassemblies are shown in Fig. 17. The top (tensile) flange generally experienced lower absolute strain than the bottom (compressive) flange, so that the neutral axis lay closer to the top flange. It is seen that the strains in both the bottom and top flanges of the steel beam at the section 100 mm away from the face of the extended end plate take values significantly higher than the yield strain of the steel. In addition, Fig. 17 shows that the strains in the top and bottom flanges of the steel beams in CJ2-CJ5 (with CLT slabs) are much higher than those of specimen SJ1 (the pure steel subassembly). It is noteworthy that the tensile/compressive strains developed in the flanges of the beams at the section 400 mm away from the face of the column, for all subassemblies remained well below the yield limit of the 310UB404.4 profile during the tests.

200 150

CJ2 100

CJ3 50

Yield strain

0 0

2000

6000

4000

Strain (με) Fig. 15. Tensile strain in spline steel plates at mid-span of specimens CJ2 and CJ3.

300 3.5. Strains in extended end plates

Load (KN)

250 The load versus strains in the extended end plates for all subassemblies are shown in Fig. 18. The yield strain of the extended end plates calculated with respect to the yield strength of the end plates is also shown in Fig. 18. The similar behaviour of all specimens (including the pure steel and STC specimens) in terms of the strain in the extended end plates is evident from Fig. 18. The extended end plates in the tension zone (at location of strain gauges EP2 and EP4, see Fig. 12b) sustained high level of strain/ stress well above the yield strength of the end plates and yielding of end plates took place at around 60% of the peak loading capacity of the STC subassemblies.

200 150

100

CJ4

50

CJ5

Yield strain

0 0

5000

10000 15000 Strain (με)

20000

25000

Fig. 16. Tensile strain in the rods used for connecting the CLT slabs in specimens CJ4 & CJ5.

(peak) load, but tensile strain in the connecting rods for specimens CJ4 remained in the elastic range, owing to premature plug shear and crushing of the CLT panels around the anchorage zone of the rod (Fig. 14), that in turn lowered the bending moment capacity and rotation capacity of the joint. Accordingly, it is concluded that the size of the connecting rods can significantly influence the structural performance of the STC connection with extended end plate and the size of connecting rods must be carefully determined to proportionate with the failure of the CLT panels.

3.6. Load-slip and load-gap opening The relative displacement/slip between the timber slab and the girder at the supported end of the STC subassemblies is shown in Fig. 19 and the maximum slip smax values are given in Table 3. The ultimate slip was within the range of 1.7–5.4 mm and the maximum and minimum slip belonged to CJ2 and CJ5, respectively. It is also shown that in the first stage of the response (up to 50 kN load), the slip for all specimens is almost zero, showing near to full composite action in the STC system was achieved by coach screw shear connectors. Despite using the same size coach screws that provided identical degree of shear connection in all STC specimens, the load-slip responses of the STC subassemblies were different.

SJ1 CJ2 CJ3 CJ4 CJ5 Bottom Flange (Compression)

200 150 100 50

0 -35000 -25000 -15000 -5000

300 Yield Strain 250

Load (KN)

Load (KN)

300 Yield Strain Yield Strain 250

Top Flange (Tension)

Strain (με)

5000

15000

SJ1 CJ2 CJ3 CJ4 CJ5

Yield Strain

200 150 100

Bottom Flange 50 (Compression) 0 -3500 -1500

Top Flange (Tension) 500

2500

Strain (με)

Fig. 17. Load-strain in flanges of the steel beam at sections (a) 100 mm and (b) 400 mm away from the face of the steel column.

647

A. Ataei et al. / Construction and Building Materials 226 (2019) 636–650

300

300

250

250

200

200

150

Load (KN)

Load (KN)

Fig. 18. Load versus strain in the extended end plate of the tested subassemblies, for strain gauge (a) EP1, (b) EP2, (c) EP3 and (d) EP4.

CJ2 CJ3 CJ4 CJ5

100 50 2

4

100 50 0

0 0

CJ2 CJ3 CJ4 CJ5

150

6

0

Fig. 19. Load-end slip response of the steel-CLT beam-to-column joints with extended end plates.

In addition, the ultimate slip between the steel beam and CLT panel observed for the composite joint tests is different. Specimens CJ4 and CJ5 (with HSS rod connections) exhibited lower slips at the end of the test than those of specimens CJ2 and CJ3 (with surface steel spline joints). This behaviour should be attributed to the lower stiffness of the timber slab-to-timber slab connections in CJ4 and CJ5 that translates the relative displacement between the slab and beam to the gap opening (see Fig. 19) instead of the slips at the supported ends of the STC beams. The comparison of the load-slip curves for subassemblies CJ2 and CJ3 (differed in the cross section and configuration of the steel plate of the surface spline CLT-to-CLT connection) are also provided in Fig. 19. It is seen that increasing the cross- sectional area of the surface steel spline joint (between the two CLT slab panels) can reduce the maximum slip at the supported end of the subassemblies. Similar results were also observed for specimens CJ4 and CJ5, where the specimen CJ4 (with 4 HSS 20 mm

10

20

30

40

Gap openingn (mm)

Slip (mm)

Fig. 20. Load versus opening of the gap between the two juxtaposed CLT slabs.

rods) exhibited lower final slip at the free end of the composite beam. The load versus opening displacement (between the two juxtaposed CLT panels) for the STC joints are shown in Fig. 20. Except for specimen CJ4, separation of the two juxtaposed timber panels started at a load around 50 kN and monotonically increased until failure. However, the separation of the CLT panels in subassembly CJ4 started at a load of 150 kN. This could be attributed to the overtightening of the nuts and excessive post-tensioning force induced in the connecting threaded rods that in turn led to premature crushing of the CLT panels in the anchorage zone (Fig. 14). The final gap opening between the two CLT panels recorded for the STC subassemblies was within the range of 16.4–32.5 mm Fig. 20. 3.7. Failure modes The failure modes of specimens which was mainly associated with either compressive buckling of the beam flange and plastic

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Fig. 21. Buckling of the beam flange and plastic deformations in the extended end plate.

deformation in the extended end plate or crushing and plug shear failure in CLT panel (only observed in CJ4) are shown in Fig. 21. A short description of the failure modes for all specimens are also provided in Table 3. All subassemblies were dismantled at the end of the push-down tests and the M24 bolts connecting the extended end plate to the column were visually inspected. The M24 bolts showed no sign of fracture, but the bolts in the tensile zone of the end plate had some plastic deformation. All semirigid steel-CLT beam-to-column connections exhibited ductile behaviour and complied with the design requirements of EC3 [42] and EC4 [43] in terms of initial stiffness, flexural resistance and rotation capacity. The test results revealed that the coach screw shear connectors have not fractured and/or sustained major plastic deformation in any of the beam-to-column subassemblies.

connections to elucidate the rotational stiffness, bending moment resistance and rotation capacity of the novel STC joints subjected to hogging bending moments, and to simulate the behaviour of an internal joint in a semi-rigid steel-timber hybrid frame. In this system, the CLT panels are attached compositely to the steel girders using coach screws and the steel-CLT composite beams are connected to the steel columns using bolted extended end plates. In addition, one specimen without a CLT slab was constructed and tested as a control specimen to assess the influence of the CLT panels on the performance of the joint. The structural behaviour of this type of joint was assessed in terms of its rotational stiffness, bending moment resistance, rotation capacity and failure modes. The main variable in the experimental program was the type of the steel connection across the steel column (surface spline with plate and/or connecting threaded rods). Based on the test results, the following conclusions were drawn.

4. Conclusion and future studies This paper deals with an experimental study of the cruciform steel-CLT beam-to-column subassemblies with extended end plate

 All tested beam-to-column joints exhibited significant nonlinearity, have sound rotation and moment capacities and failed in a ductile manner.

A. Ataei et al. / Construction and Building Materials 226 (2019) 636–650

 The composite action obtained by using a CLT panel and screw shear connectors has a significant effect on the structural behaviour of the joints.  Vertical slip at the interface between the extended end plate and the steel column flanges was not observed.  The extended end plate experienced significant deformation at the region adjacent to the beam tension flange.  Adding timber panels to the bare steel joints has a significant effect on the initial stiffness and moment capacity of the beam-to-column joints and thereby significantly improves these parameters.  The type of connection between the two CLT panels over the column has a major impact on the performance of the composite joints from structural design view point. Among the connections considered, the CLT-to-CLT spline joint with surface steel plate provided for the highest bending moment capacity and second highest stiffness of the beam-to-column connections CJ2.  All tested joints with timber panels can provide a higher rotation capacity than the ductility requirements of EC3 and EC4, and so plastic analysis and design can be utilised.  The size of the threaded rods must be determined to be proportionate with the local buckling load of the compressive flange of the steel beam and the crushing or plug shear failure load of the CLT panels.  Limited number of specimens have been tested in this study, and thereby the reliability and repeatability of the reported results require further investigation. However, the results of previous tests conducted on two identical STC subassemblies with flush end plate and continuous CLT slabs [44] have revealed that variability in the mechanical properties of CLT (even in an extrem case in which the CLT slab is continuous and thereby tensile failure of CLT governs the failure mode/load of the joints) has minor (less than 7%) influence on the ultimate bonding moment capacity and stiffness of the STC beam-tocolumn connections [44]. To further address the variability of structural behaviour of STC beam-to-column connections, in future studies, detailed nonlinear 3D finite element models of the STC subassemblies will be built, analysed and validated against the experimental results available in the literature. The validated (calibrated) FE models will be used as a tool to evaluate the sensitivity of the results with respect to the variability in the mechanical properties of the materials (e.g. timber) and construction tolerances. Declaration of Competing Interest None. Acknowledgements The work reported in this paper was funded by a Discovery Project (DP160104092) awarded to the second and third authors by the Australian Research Council. References [1] N.D. Brown, D. Anderson, Structural properties of composite major axis end plate connections, J. Constr. Steel Res. 57 (3) (2001) 327–349. [2] F. Fu, D. Lam, Experimental study on semi-rigid composite joints with steel beams and precast hollowcore slabs, J. Constr. Steel Res. 62 (8) (2006) 771– 782. [3] B. Gil, R. Goñi, E. Bayo, Experimental and numerical validation of a new design for three-dimensional semi-rigid composite joints, Eng. Struct. 48 (3) (2013) 55–69. [4] H.Y. Loh, B. Uy, M.A. Bradford, The effects of partial shear connection in composite flush end plate joints Part I — experimental study, J. Constr. Steel Res. 62 (4) (2006) 378–390.

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