Flexural performance of steel-reinforced engineered cementitious composites with different reinforcing ratios and steel types

Construction and Building Materials 231 (2020) 117159

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Flexural performance of steel-reinforced engineered cementitious composites with different reinforcing ratios and steel types Yi Shao ⇑, Sarah L. Billington Dept. of Civil and Environmental Engineering, Stanford University, Stanford, CA 94305, United States

h i g h l i g h t s
Reinforced ECC beams fail either after crack localization or gradual strain hardening of steel.
Two flexural failure paths represent different load-reduction mechanisms.
Reinforcement type and reinforcing ratio affect the flexural failure path of reinforced ECC.
A maximum compressive strain of at least 1.2%–3.0% is sustained by ECC before crushing.
A recently-developed method predicts the flexural strength of beams with an error less than 5%.

a r t i c l e

i n f o

Article history: Received 26 April 2019 Received in revised form 5 September 2019 Accepted 4 October 2019

Keywords: Engineered cementitious composite High strength steel Flexural failure path Flexural strength prediction Crack localization Maximum compressive strain

a b s t r a c t Reinforced engineered cementitious composites (ECC) flexural members fail either after a dominant crack forms (i.e., crack localization) or after gradual strain hardening of reinforcing steel. These two failure paths represent distinct ductility ranges and load reduction mechanisms that have not been fully characterized. This study experimentally investigates the two failure paths of flexural members with different reinforcing ratios (0.53%–2.10%), two types of reinforcing steel (A615 Grade 60 and A1035 Grade 100), and under monotonic and cyclic loading conditions. A total of twelve simply-supported beams are tested including two conventional concrete beams for baseline comparisons. Based on the experimental results, a recently-developed flexural strength prediction method is validated. Results show that (1) the two failure paths are affected by the reinforcing ratio and steel type, (2) ECC sustains a maximum compressive strain that is larger than 1.0% before crushing, (3) ECC effectively restrains crack width opening for flexural members using high strength (Grade 100) steel reinforcement, and (4) a recently proposed flexural strength prediction method is able to predict the experimentally observed flexural strength of the beams with an error less than 5%. Ó 2019 Elsevier Ltd. All rights reserved.

1. Introduction Compared to conventional concrete, Engineered Cementitious Composites (ECC) shows higher composite ductility in both tension and compression [1]. This higher composite ductility makes ECC a promising material for structures under extreme loadings, such as earthquakes [2,3]. However, there are several unique aspects of reinforced ECC compared to reinforced concrete that are still not well understood. Reinforced ECC flexural members have been observed to fail either after a dominant crack forms (i.e., crack localization) or after gradual strain hardening of reinforcing the steel [4]. These two failure paths represent distinct ductility ranges ⇑ Corresponding author. E-mail addresses: [email protected] (Y. Shao), [email protected] (S.L. Billington). https://doi.org/10.1016/j.conbuildmat.2019.117159 0950-0618/Ó 2019 Elsevier Ltd. All rights reserved.

and load reduction mechanisms that have not been fully characterized. In addition, the high spalling resistance of ECC has been experimentally observed for reinforced ECC [1,5–7], the maximum compressive strain that ECC can sustain before crushing remains largely unknown and is needed for flexural strength predictions. The primary objective of this study is to provide a systematic examination and characterization of the two failure paths of reinforced ECC flexural members and their load-reduction mechanisms to further develop and validate flexural strength prediction methods that may be used for design. The flexural failure paths and load-reduction mechanisms of reinforced ECC beams with different reinforcing ratios (0.53%–2.1%), different types of reinforcing steel (A615 Grade 60 and A1035 Grade 100) and different loading conditions (monotonic and cyclic) is examined here. Additional goals of this study are to examine the crack opening of Grade 100 steel reinforced ECC flexural elements under representative service

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loads and the cyclic performance of Grade 100 steel reinforced ECC structural members. Ten reinforced ECC beams and two reinforced concrete beams are tested. Maximum crack widths, load-drift response, moment–curvature response, reinforcing steel strain, and maximum compressive strain are measured and discussed. A recently proposed flexural design method is evaluated using the experimental results.

concrete structures [16,17], which limits the use of Grade 100 steel in structures with a seismic design category of D or higher [14]. Thus, cyclic testing of Grade 100 steel reinforced structural components is needed to expand the knowledge of its cyclic performance. 3. Experimental program 3.1. Materials

2. Background Recent tension-stiffening experiments of reinforced ECC demonstrate that, under tension, ECC initially deforms compatibly with reinforcing steel and provides additional tensile capacity to the reinforced specimen [8–10]. When ECC begins to lose fiberbridging capacity (i.e., crack localization), the ECC tensionstiffening specimens also lose load-carrying capacity if the ECC fiber-bridging capacity is larger than the remaining steel hardening capacity at the point when crack localization begins [8,9]. These specimens fail at just 2–4% specimen strain, indicating that reinforcing strains of 9–20% were reached within the cracked regions of these tension-stiffening specimens. On the other hand, if the ECC fiber-bridging capacity is smaller than the remaining steel hardening capacity at the point of crack localization, the reinforcing steel will continue to carry load to larger deformations through hardening of the reinforcing steel. While ECC can maintain tensile strength to relatively large strains (e.g., 1.0%) compared to conventional concrete, this strain capacity is smaller than the ultimate strain of reinforcing steel (e.g., 9–20%), and understanding what behavior will dominate after crack localization is essential for understanding the strength and ductility of reinforced ECC flexural members. Similar to the two failure modes of reinforced ECC tension stiffening members, two dominant flexural failure paths have been identified for reinforced ECC flexural members [4]: (1) failure may occur after crack localization (i.e., the formation of a single, dominant crack), and the load capacity reduces because of the loss of fiber-bridging capacity that cannot be compensated for by the hardening of longitudinal reinforcing steel [5,11,12], or (2) failure may occur after crack localization, whereby the beam achieves a higher load capacity through gradual strain hardening of longitudinal reinforcing steel [5,11,13]. These two failure paths, termed ‘failure after crack localization’ and ‘failure after gradual strain hardening’, represent distinct load-reduction mechanisms and ductility ranges. In studying failure mechanisms of reinforced ECC, two types of reinforcing steel are examined here: A615 Grade 60 and A1035 Grade 100 steel. While Grade 60 steel is widely adopted in practice, Grade 100 steel has been proposed as reinforcement in traditional reinforced concrete structures [14] because: (1) it requires less reinforcing steel area than traditional Grade 60 steel and thus reduces the costs associated with handling; (2) the reduced volume of steel helps avoid steel congestion, and; (3) it shows enhanced corrosion resistance. However, higher steel stresses under service loads may lead to large crack widths and require longer development lengths [14]. As ECC exhibits good cracking resistance through fine, multiple cracking [1] and enhanced bond strength with reinforcement [15], ECC is potentially an ideal matrix material for Grade 100 steel. To date, few experiments have been dedicated to the cyclic behavior of Grade 100 steel reinforced

Table 1 lists the mixture design of ECC and concrete. The ECC mixture was chosen for its suitability for large-scale applications [18] and to compare the results to similar work using the same mixture [9,15,19]. The ECC mixture uses 8-mm-long polyvinyl alcohol (PVA) fibers with a volume fraction of 2%, Type II/V cement, class F fly ash, #90 silica sand, a viscosity modifying admixture (VMA), and a high range water-reducing admixture (HRWR). The concrete was designed to have similar compressive strength to the ECC mixture. This concrete mixture contained Type II/V cement, class F fly ash, coarse aggregate with a maximum size of 9.5 mm, and sand with a fineness modulus of 2.6. The ECC was mixed in a horizontal pan mortar mixer, and the concrete was mixed in a tilting drum mixer. Separate batches were cast for each reinforced beam due to the limited capacity of the mixer. Three ECC cylinders with a diameter of 100 mm and a height of 200 mm and three unreinforced ECC beams with a cross-section of 75 mm  75 mm and length of 275 mm were cast with each reinforced ECC beam, and three concrete cylinders with a diameter of 100 mm and a height of 200 mm were cast with each reinforced concrete beam. The beams and cylinders were removed from the molds 24 h after being cast, and were then moist cured for 7 days. After moist curing, the beams and cylinders were air cured until the test date of 70 ± 2 days. The cylinders were tested in compression according to ASTM C39-16b. Compressive strain prior to the peak load was measured by a digital image correlation (DIC) system over the middle 100 mm of the cylinder height. After the peak load was reached, surface damage disturbed the DIC system, and the post-peak deformation of the cylinders was obtained from the actuator displacement with a correction method first proposed by Mansur et al. [20]. This method accounts for machine flexibility and specimen end conditions. The average compressive strengths for the concrete and the ECC were 49.6 MPa and 49.9 MPa with standard deviations of 4.0 MPa and 3.0 MPa, respectively. The concrete compressive stress–strain curve (Fig. 1a) was comprised of a linear ascending portion leading into a nonlinear ascending portion up to the peak strength. After the peak was reached, a small load-drop occurred followed by an unstable load drop accompanied by diagonal shear cracks and severe spalling. After the unstable load-drop, no residual compressive strength was maintained in the concrete cylinders. Two failure modes were observed in the ECC compression cylinder after reaching peak strength (Fig. 1a): (1) shear sliding: a diagonal shear crack plane formed with an unstable load drop, after which fiber-bridging maintained the integrity of the cylinder and shear sliding across shear plane provided nominal residual compressive strength; (2) lateral expansion: compressive strength was gradually lost with lateral expansion of the cylinder, and the fiber-bridging prevented spalling of the cylinder. Shear sliding with an unstable load drop commonly occurs in both ECC and concrete

Table 1 Mixture proportion for 1 m3 of ECC and Concrete. Matrix

Cement (kg)

Fly Ash (kg)

Coarse Aggregate (kg)

Fine Aggregate (kg)

Water (kg)

VMA (% wt. cement)

HRWR (% wt. cement)

PVA (% vol.)

ECC Concrete

547 590

656 147

0 671

438 622

312 241

0.11 0

0.50 0

2.0 0

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Fig. 1. ECC material behavior: (a) compressive stress-strain relationship from experiment; (b) load-deflection response of unreinforced ECC beam from experiment and numerical simulation; (c) ECC tensile stress strain relationship from numerical simulation. ‘h’: measurement stopped.

cylinder tests because, once the cylinder strain exceeds the softening strain, no surrounding material continues supporting the compressive force. The strain energy in the testing machine is released and the specimen may fail explosively [21]. In ECC flexural members, once the compressive region (i.e., top of the beam) enters the softening branch, the surrounding material remains intact and can sustain compression load while fiber-bridging limits the spalling of the softened material. Explosive failure is therefore avoided in the compression zone as discussed in Section 4.4. Thus, cylinder lateral expansion is believed to better represent ECC behavior in the compressive zone of flexural members. The unreinforced ECC beams were tested in third-point bending with a total span of 225 mm. The average equivalent flexural strength of the ECC was 10.3 MPa with a standard deviation of 1.5 MPa. Two-dimensional nonlinear finite element simulations were conducted on the unreinforced ECC beams to estimate the direct tension performance, and from this, the ECC was estimated to have an effective uniaxial tensile strength of 4.0 MPa, which is consistent with a previous study with the same mix design [22]. Fig. 1b compares the load–deflection response of the unreinforced beam from the experiment and the numerical simulation. Fig. 1c shows the estimated ECC tensile stress–strain relationship. A trilinear tensile curve with a plateau was adopted in Fig. 1c, and the descending branch is considered using tensile fracture energy, Gf [23]. Although the ECC often shows uniaxial tensile strainhardening behavior, neglecting the hardening (and assuming an elastic-perfectly plastic response as shown in Fig. 1c) is sufficient for structural-level analysis and is a common assumption in literature (e.g., [23–25]) and standards (e.g., [26]). A615 Grade 60 normal strength steel (denoted as ‘NSS’ hereafter) and A1035 Grade 100 high strength steel (denoted as ‘HSS’ hereafter) were used for longitudinal reinforcement. Three pieces of steel reinforcement for each type of steel were tested to obtain the tensile stress–strain relationship. An extensometer with a gauge length of 50 mm was used to measure the steel strain and was removed before steel fracture to protect the equipment. Fig. 2 shows a representative stress–strain relationship from each of the sizes of NSS and HSS and the geometric and material property details are given in Table 2. Only No.4 NSS reinforcement had a well-defined yield plateau. Shear stirrups were A615 Grade 60 steel with a diameter of 9.5 mm and a manufacturer-reported yield strength of 510 MPa. 3.2. Specimen design and test setup Fig. 3 shows the details of the specimen design and test setup. Each beam was 2000 mm in length, 150 mm wide, and 220 mm deep. A shear span of 800 mm was chosen to guarantee flexural failure and fit the capacity of the testing machine. The shear stirrup

Fig. 2. Stress-strain response of reinforcing steel in tension.

spacing was 90 mm, which was ½ d (d = the distance from the extreme compression fiber to the centroid of the bottom reinforcement). Table 3 gives the details and the predicted nominal moment capacity of all of the tested beams. The prediction method is presented in Section 5.2. The naming convention indicates the steel type, matrix material, reinforcing ratio (q), and loading type. The beams were subjected to four-point bending at a quasistatic displacement rate of 0.097 mm/s, which was controlled by the actuator of the 245 kN testing machine (MTS Systems, Eden Prairie, Minnesota) (Fig. 3). The constant moment region of one side of the beam was monitored by a digital image correlation (DIC) system. Measurements of curvature, maximum compressive strain, and maximum crack width were monitored during each experiment by creating extensometers within the DIC system. In the constant moment region, the length of extensometers was 200 mm (for curvature and compressive strain measurements). The maximum crack width was measured by a virtual extensometer with a length of about 10 mm in the DIC system that bridged the bottom of the localized crack. Mid-span deflection was obtained by averaging the reading from two string pots. Load was recorded by the testing machine load cell. One Omega KFH strain gauge with a strain measuring capacity of 2% was attached to the middle point of each longitudinal reinforcing bar as shown in Fig. 3. 3.3. Loading protocol For each pair of the beams using the same steel type, matrix material, and reinforcing ratio, one beam was tested under

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Table 2 Reinforcing Steel Properties.

NSS NSS HSS HSS HSS

Bar size

Diameter [mm]

fy [MPa]

fu [MPa]

Elongation* [%]

No.4 No.6 No.3 No.4 No.5

12.7 19.1 9.5 12.7 15.9

455 422 846 734 743

655 640 1285 1152 1183

15.6 15.6 8.0 9.0 9.0

Note: fy = yielding strength determined by 0.2% offset method; fu = ultimate strength; *=provided by manufacturer.

Fig. 3. Specimen design and test setup.

Table 3 Flexural Test Specimens. Name Convention

Steel Type

Matrix

q

Loading

Bar Size

Predicted Moment Capacity [kN-m]

HSS-CON-0.53-M HSS-CON-0.53-C HSS-ECC-0.53-M HSS-ECC-0.53-C HSS-ECC-0.96-M HSS-ECC-0.96-C HSS-ECC-1.48-M HSS-ECC-1.48-C NSS-ECC-0.96-M NSS-ECC-0.96-C NSS-ECC-2.10-M NSS-ECC-2.10-C

HSS HSS HSS HSS HSS HSS HSS HSS NSS NSS NSS NSS

Concrete Concrete ECC ECC ECC ECC ECC ECC ECC ECC ECC ECC

0.53% 0.53% 0.53% 0.53% 0.96% 0.96% 1.48% 1.48% 0.96% 0.96% 2.10% 2.10%

Monotonic Cyclic Monotonic Cyclic Monotonic Cyclic Monotonic Cyclic Monotonic Cyclic Monotonic Cyclic

No. No. No. No. No. No. No. No. No. No. No. No.

28.5 28.5 36.4 36.4 48.7 48.7 70.2 70.2 27.6 27.6 53.1 53.1

monotonic load and the second was tested under cyclic load. The cyclic loading protocol was chosen based on FEMA 461 [27], which is comprised of cycles of step-wise increasing deformation amplitudes (Fig. 4). At each step, two cycles were completed. The amplitude of each step was 1.4 times the amplitude of the previous step

3 3 3 3 4 4 5 5 4 4 6 6

until reaching the target amplitude Dm , which was the deformation level at the peak load of the corresponding monotonic specimen. If the cyclic specimen did not fail at Dm , then testing continued with the amplitude being increased to 1.3 times the previous step until failure. All monotonic and cyclic beams were tested until reinforcement fracture with the exception of NSS-ECC-0.96-C and NSS-ECC2.10-C where steel fracture did not occur before the displacement range of the test setup was reached.

4. Results 4.1. Monotonic cracking behavior and load-displacement response

Fig. 4. Cyclic loading protocol.

Fig. 5 presents the crack patterns of the six monotonic beams just before fracture of the reinforcement with the location of reinforcement fracture denoted by an ‘X’. The drift (defined as the ratio between mid-span deflection and shear span length) at the reinforcement fracture point is shown at the upper-right corner of each beam. Compared to the reinforced concrete beam, the reinforced ECC beams generally exhibited a denser cracking pattern as well as more longitudinal splitting cracks. The ECC beams of similar strength (i.e., HSS-ECC-0.53-M and NSS-ECC-0.96-M, HSS-ECC-0.96-M and NSS-ECC-2.10-M) show a similar flexural cracking at failure with the beams reinforced with HSS showing

Y. Shao, S.L. Billington / Construction and Building Materials 231 (2020) 117159

Fig. 5. Crack pattern before longitudinal steel fracture.

more longitudinal splitting cracks, which is attributed to the higher stress developed in the HSS bars. The beams with higher strength exhibited more dense flexural and shear cracking as expected because the higher load led to larger flexural and shear stresses along the beam, which in turn activated the multiple-cracking phenomenon in the ECC [10,28]. The majority of the cracks in the ECC beams formed before the longitudinal steel began yielding and, up to this point, they had a maximum crack width below 0.3 mm. After steel yielding, one or two cracks began opening unstably, which is referred to as ‘crack localization’. The crack localization point was assumed to occur when the maximum crack width reached 0.4 mm, which is when fiber-bridging fails and triggers load-reduction in unreinforced ECC beams. Fig. 6 shows the load–displacement response of all monotonically tested beams. In Fig. 6 crack localization is denoted on each response with an ‘O’ and reinforcement fracture is denoted with an ‘X’. Displacement is expressed as the specimen drift, which is the ratio between mid-span deflection (D) and shear span length. Despite the lower elastic modulus of plain ECC (Fig. 1a), the reinforced ECC beams showed a higher post-cracking stiffness than

Fig. 6. Load-displacement response. ‘X’: longitudinal reinforcement fracture; ‘O’: crack localization.

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the reinforced concrete beams (Fig. 6) because the fiber-bridging in the tension zone increased the section stiffness, which is commonly observed in fiber-reinforced concrete specimens (e.g., [9,29–31]). Failure of the reinforced concrete beam was typical for an under-reinforced beam and consisted of crushing of the concrete after steel yielding. Given the low reinforcing ratio for the concrete beam, the reinforcement fractured soon after the load loss from crushing of the concrete. The reinforced ECC beams showed two distinct failure paths after crack localization as seen in Fig. 6. The failure path termed ‘failure after crack localization’ was observed for Specimens HSS-ECC-0.53-M and HSS-ECC-0.96-M. The remainder of the specimens followed the failure path termed ’failure after gradual strain hardening’. For HSS-ECC-0.53-M and HSS-ECC-0.96-M, their load capacity slightly increased after crack localization with rapid steel strain hardening. The beams then rapidly lost their load-carrying capacity due to the loss of fiber-bridging capacity and the reinforcing steel fractured. A failure drift of 4.5–5.3% was observed for these two beams. No visual compressive damage was observed when load began dropping. At the peak load of these two beams, the localized crack was restrained at the bottom of the beam extending up less than 25% of the section height, and the maximum crack width at the bottom face was less than 2.5 mm, which indicates that fiber-bridging was still being maintained across the tensile zone. For specimens HSS-ECC-1.48-M, NSS-ECC-0.96-M, and NSSECC-2.10-M, their load capacity after crack localization gradually increased with the strain hardening of the longitudinal reinforcing steel. Final failure of these three beams was initiated by crushing of the ECC followed by fracture of the longitudinal steel. Failure drifts of 8.9–15.5% were observed for these beams. At the peak load of each of these beams, the localized crack extended up to 55%–65% of the section height. The maximum crack width at the bottom face exceeded 12 mm, indicating that fiber-bridging was no longer happening in most of the tensile zone. The larger crack widths and larger deformations at the peak load for specimens that fail after gradual strain hardening allow for more warning of impending failure than specimens that fail after crack localization. 4.2. Ductility of the monotonically-loaded beams With lower reinforcing ratios of HSS (i.e., up to 0.96% in this study), reinforced ECC beams showed a smaller drift capacity than the reinforced concrete specimen (i.e., HSS-CON-0.53-M) despite the higher composite ductility of ECC (Fig. 6). HSS reinforced ECC beams with low reinforcing ratios lost load-capacity not from ECC crushing but rather from fiber-bridging failure, which occurred at a small curvature. After crack localization, rapid strain hardening of the reinforcing steel occurred (Section 4.3), attempting to compensate for the load capacity loss from fiber-bridging failure. Due to the high bond strength between ECC and the reinforcement [15], steel hardening was restrained to a small region near the localized crack and steel fractured early at small drift, which is also observed from flexural tests [5,23] and tension-stiffening tests [9] of reinforced ECC as discussed in Section 2. Similar observations have been made for steel reinforced ultra-high performance concrete (UHPC) beams: adding fibers to UHPC enhanced the material tensile performance, but also led to failure at a lower drift after the point of crack localization [32,33]. For HSS and NSS reinforced ECC beams, increasing the reinforcing ratio increased the drift capacity of beams. This trend is consistent with previous studies on reinforced ECC [5] and reinforced UHPC [34,35]. Especially for the HSS reinforced specimens, increasing the reinforcing ratio from 0.96% to 1.48% changed the failure path from failure after crack localization to failure after gradual strain hardening thus increasing the drift capacity by 68%. This change of failure path with increasing reinforcing ratio was due

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to the fact that with a higher reinforcing ratio, the reinforcing steel’s hardening capacity was able to compensate for the load capacity reduction from the loss of fiber-bridging capacity after crack localization. This impact of reinforcing ratio on failure path is also observed in reinforced UHPC beams [35]. For the reinforced ECC beams with a reinforcing ratio of 0.96%, changing the reinforcing steel from HSS to NSS increased the failure drift by 130%. This increase in failure drift occurred because the failure path changed due to the larger steel strain-hardening capacity remaining in the NSS reinforced ECC beam at the crack localization point relative to that of the HSS reinforced beam. At crack localization, the contribution of fiber-bridging to the nominal moment capacity was estimated to be 11.2 kN-m for beam HSSECC-0.96-M and 11.7 kN-m for beam NSS-ECC-0.96-M, based on the measured crack heights and ECC effective tensile strength. The measured steel strain in HSS-ECC-0.96-M was 1.5%, which implies a remaining increase in moment capacity of 5.2 kN-m from hardening of steel. The measured steel strain in NSS-ECC-0.96-M at crack localization was 0.2%, which implies a remaining increase in moment capacity of 8.9 kN-m from hardening of steel. As fiberbridging capacity was gradually lost in both beams after crack localization, beam NSS-ECC-0.96-M having the larger remaining hardening capacity was able to reach a higher ductility than beam HSS-ECC-0.96-M.

The importance of designing for the appropriate failure path can be assessed by examining the difference in response between HSSECC-0.96-M and NSS-ECC-2.10-M. HSS-ECC-0.96-M had similar load capacity as NSS-ECC-2.10-M (Fig. 6). However, NSS-ECC2.10-M had a drift capacity that is 193% larger than the drift capacity of HSS-ECC-0.96-M. While HSS-ECC-0.96-M failed after crack localization at a drift of 5.3%, NSS-ECC-2.10-M with its higher reinforcing ratio had larger steel hardening capacity than NSS-ECC0.96-M, and thus failed after gradual strain hardening at a drift of 15.5%. This comparison between HSS-ECC-0.96-M and NSSECC-2.10-M implies that, when designing reinforced ECC flexural members, the desired failure path as well as the desired flexural strength should be designed for to provide sufficient deformation capacity and thus warning prior to failure. 4.3. Reinforcement strain in monotonic beams Fig. 7 presents the reinforcement strain history along with the load–displacement response of each monotonic beam. The steel strain plotted is the average of the reading from two strain gauges at the middle of the tension steel until the measuring range (2%) is exceeded. For NSS-ECC-0.96-M, one of the two strain gauges failed earlier than the other, the reported steel strain from the average and single strain gauge reading is differentiated by line type. The

Fig. 7. Bottom reinforcement strain history. ‘O’: crack localization; ‘X’: longitudinal reinforcement fracture; ‘h’: strain gauges exceed measuring capacity.

Y. Shao, S.L. Billington / Construction and Building Materials 231 (2020) 117159

difference of the readings from the two strain gauges was less than 5%. At crack localization, the measured steel strain ranged from 0.9% to 1.5% for the HSS reinforced beams, and ranged from 0.2% to 0.6% for the NSS reinforced beams. These strains indicate that the reinforcing steel in all of the beams has yielded when crack localization occurred. At the peak load for the specimens that failed after crack localization (i.e., HSS-ECC-0.53-M and HSS-ECC-0.96M), the measured steel strain was 1.6%, which indicates a small amount of strain hardening of the steel. After the measured steel strain exceeded the strain gage limit (i.e., 2%) (Fig. 7), the specimens that failed after gradual strain hardening underwent another 2.2–7.0% drift before reaching the peak load. These strains and further deformation indicate that a significant amount of steel strain hardening has occurred at peak load for the beams that failed after gradual strain hardening (refer to Fig. 2 for steel stress strain response). 4.4. Monotonic compressive zone behavior Fig. 8 shows the maximum compressive strain history of the six monotonic beams. The ‘h’ represents the point where surface damage prohibited the use of the DIC system and the ‘X’ represents the point when the longitudinal reinforcement fractured. When the maximum compressive strain reached 0.0048, HSS-CON-0.53-M

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began crushing and losing load capacity as would be expected with reinforced concrete. At the peak load for the reinforced ECC specimens failing after crack localization, the measured maximum compressive strain was 0.004 for both HSS-ECC-0.53-M and HSSECC-0.96-M; no compressive damage was observed at the peak load for these two beams. From cylinder compression tests, the compressive strain at peak stress was measured to be between 0.004 and 0.005, which means, at the peak load, the compressive zone in the two beams had not entered the softening state. On the other hand, at the peak load for specimens failing after gradual strain hardening, the strain at the extreme compressive fiber was 0.012, 0.028, and 0.030 for NSS-ECC-0.96-M, NSS-ECC-2.10-M, and HSS-ECC-1.48-M respectively. 95% of that peak load was then maintained for compressive strains beyond 0.03 for all three beams (Fig. 8). At a compressive strain of 0.03, the extreme compressive fiber likely sustained approximately 30% of its original compressive strength (Fig. 1a) due to the high-spalling resistance of ECC, which is enabled by fiber-bridging. After softening initiated in the compressive zone, the centroid of compression force shifted down towards the neutral axis, which resulted in a smaller moment arm and potentially smaller moment capacity. The beams sustained a higher load until a relatively large compressive strain (up to 0.03) because (1) fiber-bridging prevents severe spalling and rapid compression force loss, which is commonly observed in concrete (e.g., HSS-CON-0.53-M in this study), and; (2) the

Fig. 8. Maximum compressive strain history. ‘X’: longitudinal reinforcement fracture; ‘h’: Surface strain measurement stopped.

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continuous hardening of the tension reinforcement along the bottom of the beam provided additional load capacity, which is shown in Fig. 7 and further discussed in Section 5.2.2. After reaching peak load, HSS-CON-0.53-M exhibited significant spalling while the reinforced ECC beams maintained the integrity of the concrete cover (Fig. 9) which is consistent with previous studies [7,31]. 4.5. Impact of cyclic loading Fig. 10 compares the load–displacement response of the beams subjected to monotonic loading and to cyclic loading. A ‘h’ represents the point where the test was stopped due to the deformation capacity of the test setup being exceeded and an ‘X’ represents the point where longitudinal reinforcement fracture occurred. At the end of the tests for NSS-ECC-0.96-C and NSS-ECC-2.10-C, severe cracking with maximum crack widths larger than 10 mm had formed and led to the loss of functionality. At this stage both beams were considered to have failed. Generally, the load–displacement envelope of cyclically-loaded specimens followed a similar path to that of the monotonically-loaded specimens but with a reduced drift capacity, which is consistent with a previous study of monotonically and cyclically loaded reinforced ECC beams [5]. The flexural strength of cyclically-loaded specimens was 89%– 106% of the flexural strength of the corresponding monotonicallyloaded specimens with a mean absolute difference of 4%, which indicates that the cyclic loading had a minor influence on flexural strength. From monotonic to cyclic loading, the failure drift was reduced by 25%–33% in the reinforced ECC beams, and by 25% in the reinforced concrete beam. The cyclically-loaded specimens are expected to show a smaller drift capacity because the cyclic loading led to the low cycle fatigue of reinforcing steel and the associated earlier steel fracture [36]. In addition, load cycles may cause the steel strain to accumulate near the localized crack [5,23], which also contributes to the earlier steel fracture.

The maximum drift for the cyclically loaded HSS reinforced ECC and concrete specimens ranged from 3.3% to 6.0%. This drift capacity range exceeds the maximum drift allowed by ASCE 7–16 in a concrete structural member under seismic load (i.e., 2.5%) [37]. From Fig. 10, it is observed that the hysteretic response of the HSS reinforced ECC specimens is more pinched than that of the NSS reinforced ECC specimens. This difference in hysteretic response has also been observed between HSS and NSS reinforced concrete specimens [16,17], and is attributed to the difference in steel cyclic behavior, which needs further investigation. 5. Discussion 5.1. Maximum crack width of HSS-reinforced ECC beams For serviceability, design codes require maximum crack widths (wmax ) to be below a specified limit either explicitly [38,39] or implicitly [40], which typically ranges from 0.2 mm to 0.4 mm depending on the exposure condition. To assess the wmax of HSS reinforced ECC and HSS reinforced concrete, Table 4 lists the drift and load that corresponds to a maximum crack width (wmax ) of 0.2 mm and 0.4 mm. Similar to drifts reported in Section 4.1, drift is defined as the ratio of mid-span deflection to shear span. For HSS-CON-0.53-M, the load reached 15% and 42% of its peak load when the maximum crack width (wmax ) reached 0.2 mm and 0.4 mm, respectively. These results are in line with previous work where it has been noted that cracking in HSS-reinforced concrete specimens would be difficult to control under service loads [14,41]. In contrast, for the reinforced ECC specimens, the loads corresponding to a maximum crack width of 0.2 mm and 0.4 mm were 78%–98% of the beams’ peak loads. ECC as a matrix material is able to effectively restrain crack opening for both NSS- and HSSreinforced members and could help address the concern of large crack widths at service stress for HSS in reinforced concrete [14].

Fig. 9. Compressive zone behavior after longitudinal steel fracture.

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Fig. 10. Cyclic versus monotonic Load-displacement response; ‘X’: longitudinal steel fracture, ‘h’: Test stopped.

Table 4 Load and drift corresponding to typical service load crack widths. Specimen

HSS-CON-0.53-M HSS-ECC-0.53-M HSS-ECC-0.96-M HSS-ECC-1.48-M NSS-ECC-0.96-M NSS-ECC-2.10-M

wmax ¼ 0:2 mm

wmax ¼ 0:4 mm

Peak Load (kN)

Load (kN)

Drift (%)

Load (kN)

Drift (%)

11.6 70.5 113.2 148.8 59.2 105.8

0.3 1.8 2.6 2.6 1.2 1.4

31.3 80.8 119.3 161.4 60.4 115.6

1.0 2.4 2.8 3.5 1.3 2.0

5.2. Predicted vs. Measured flexural strength As reviewed in Shao and Billington [4], current flexural strength prediction methods for reinforced ECC beams follow an approach similar to that of traditional reinforced concrete (i.e., using a crushing strain as the limit state strain) with additional consideration of fiber-bridging in the tension zone (e.g., [26,42,43]). As demonstrated by the experimental program here, the two failure paths

75.4 88.3 121.2 176.3 68.3 134.9

represent different load-reduction mechanisms as well as tensile and compressive zone behavior, which calls for using different limit states for predicting strength. Section 5.2.1 briefly reviews a recently-developed flexural strength prediction method that accounts for the differences observed in the two failure paths [4]. Section 5.2.2 compares the predicted moment–curvature response using the proposed method to the experimentally-measured moment–curvature response.

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5.2.1. Proposed method Fig. 11 shows the strain and stress distribution for determining the flexural strength of reinforced ECC beams according to the two failure paths. For beams that fail after gradual strain hardening, the maximum flexural strength is reached when the maximum compressive strain reaches the ultimate strain ecu . As discussed in Section 4.4, measured ecu ranged from 0.01 to 0.03. Also, more than 95% of load capacity was maintained when the maximum compressive strain was 0.03. As a result, ecu was assumed as 0.03 in this method. For beams that fail after gradual strain hardening, the composite tensile strength was neglected because the localized crack has fully opened and nearly-no fiber-bridging was maintained at peak load. The strain hardening of steel is considered for beams failing after gradual strain hardening. For beams that fail after crack localization, the maximum flexural strength is reached when the localization strain of composite was reached at the tension reinforcing steel level (i.e., eloc ¼ es ¼ 0:01). At peak load, the steel yielding strength is adopted and composite tensile strength is assumed to bridge the entire tension zone. For both failure predictions, the neutral axis position is determined by force equilibrium, plane sections are assumed to remain plane and perfect bond is assumed between the HPFRCC and the reinforcing steel. The yielding point for both failure paths was determined using the same method for conventional concrete design with the only difference being that fiber-bridging was assumed to exist in the entire tension zone. ECC compressive stress–strain relationship was a modified Hognestad model [21]: a parabolic ascending branch followed by a linear descending branch with 70% compressive strength reduction when ecu ¼ 0:03, which approximates the lateral expansion failure mode from cylinder compression test in Fig. 1a. ECC tensile stress–strain relationship was a linear ascending branch with a plateau (Fig. 1c). The measured steel stress– strain relationship as shown in Fig. 2 was adopted.

5.2.2. Validation with experiments Fig. 12(a) compares the predicted moment–curvature response using the proposed flexural strength prediction method to the measured moment–curvature response for the three beams that failed after gradual strain hardening. The predicted path consists of four points that are, from left to right in Fig. 12a: (1) an initial point (i.e., moment and curvature are zero); (2) the yielding point; (3) a top fiber compressive strain of ec ¼ 0:01; and; (4) an assumed crushing with ecu ¼ 0:03. The flexural strength prediction error for these three beams was less than ±5%. When ec ¼ 0:01, the calculated bottom steel strain for these three beams ranges from 0.026 to 0.039, and 75.8% to 79.1% of the steel post-yield hardening capacity is utilized. When ec ¼ 0:03, the calculated bottom steel strain for these three beams ranges from 0.089 to 0.107, and the ultimate steel strength is reached. This change of bottom steel strain implies that with a higher compressive strain (from 0.01 to 0.03), although the top of the compression zone is softening, the load capacity still increases with continuous hardening of the bottom reinforcing steel. With further increases of ec (e.g., from 0.03 to 0.04), the load capacity is expected to drop because the bottom steel has already fully hardened and the continuous softening of top fiber will decrease the load-carrying capacity. Fig. 12(b) compares the predicted moment curvature response to the measured moment–curvature response for the two beams that failed after crack localization. The predicted path consists of four points that are, from left to right in Fig. 12b: (1) an initial point; (2) the yielding point; (3) the point of crack localization when eloc ¼ es ¼ 0:01, and; (4) the point at which the top fiber compressive strain of ec ¼ 0:01 is reached. The flexural strength prediction error for these two beams was less than ± 5%. Overall, the predicted moment–curvature response at key points captures the envelope of the experimental data. Fig. 12(b) also presents the moment–curvature prediction for HSS-CON-0.53-M using the ACI 318–14 method [40]. The predicted path consists of (1) an initial point; (2) the yield-

Fig. 11. Stress and strain distribution for predicting the flexural strength based on two failure paths.

Fig. 12. Predicted versus measured moment–curvature response for (a) failure after gradual strain hardening; (b) failure after crack localization.

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ing point, and; (3) the point of crushing when ecu ¼ 0:003. The flexural strength prediction error for concrete beam was 6%. 6. Conclusions With comparison to traditional concrete specimens, this paper examines the cracking behavior, load–displacement response, flexural failure path, reinforcing steel strain development, compressive zone behavior, impact of cyclic loading, and strength prediction methods for reinforced ECC flexural members with different reinforcing ratios and steel types. The following conclusions are reached: (1) For reinforced ECC beams that fail after crack localization, the load-reduction is triggered by the loss of fiber-bridging capacity; for reinforced ECC beams that fail after gradual strain hardening, the load-reduction is triggered by crushing. (2) The two failure paths for reinforced ECC are affected by reinforcing ratios and steel type, and represent different failure drift capacities ranging from 4.5 to 5.3% for failure after crack localization and from 8.9 to 15.5% for failure after gradual strain hardening. (3) If failing after crack localization, reinforced ECC beams show a smaller drift capacity than an equivalently reinforced concrete beam despite the higher material ductility of ECC because, compared to the load-reduction from crushing of reinforced concrete, the load-reduction from crack localization in reinforced ECC happens at a relatively small drift (or curvature). (4) A maximum compressive strain of at least 1.2%–3.0% is sustained by ECC before crushing in all of the reinforced ECC beams tested here. (5) At the peak load for all of the reinforced ECC beams that failed after crack localization, the reinforcing steel has yielded with slight strain hardening, and there is fiberbridging throughout the tension zone. At the peak load for the beams that failed after gradual strain hardening, the reinforcing steel has strain hardened to strains larger than 2.0%, and there is no fiber bridging across the dominant crack up to 55%–65% of the specimen height. (6) When designing reinforced ECC flexural members, it is proposed that engineers design for the desired failure path as well as the corresponding (and necessary) flexural strength. (7) The failure drift for the reinforced ECC specimens reduces by 25%–33% when tested cyclically compared to monotonically, and for the reinforced concrete specimens the drift capacity is reduced by 22% when tested cyclically compared to monotonically. Cyclic loading may be expected to reduce the drift capacity because of low cycle fatigue of the reinforcing steel [36] in combination with steel strain accumulation near the localized crack [5,23]. (8) Under cyclic loading, HSS steel reinforced beams fail at a drift between 3.3% and 6.0%, which meets the ASCE 7-16 code requirements. (9) The use of ECC in place of concrete could address the concern of large crack widths at service stresses for high strength steel-reinforced concrete beams. Crack widths are well controlled by the ECC due to its propensity for multiple fine cracking under tension. (10) A recently developed flexural strength prediction method is able to predict the flexural strength of all of the reinforced ECC beams tested here with an error less than 5%. The predicted moment–curvature response at key points captures the envelope of the experimental data.

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Declaration of Competing Interest The authors declare that they have no known competing financial interests or personal relationships that could have appeared to influence the work reported in this paper.

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