Hazard mitigation and strengthening of unreinforced masonry walls using composites

Composite Structures 73 (2006) 458–477 www.elsevier.com/locate/compstruct

Hazard mitigation and strengthening of unreinforced masonry walls using composites W.W. El-Dakhakhni a

a,*

, A.A. Hamid a, Z.H.R. Hakam b, M. Elgaaly

c

Department of Civil Engineering, McMaster University Centre for Effective Design of Structures, Hamilton, ON, Canada L8S 4L7 b Bechtel Power Corporation, Frederick, MD 21703, USA c Civil, Architectural and Environmental Engineering Department, Drexel University, Philadelphia, PA 19104, USA Available online 9 April 2005

Abstract An experimental investigation was conducted to study the behavior of unreinforced masonry (URM) walls retrofitted with composite laminates. The first testing phase included testing 24 URM assemblages under different stress conditions present in masonry walls. Tests included prisms loaded in compression normal and parallel to bed joints, diagonal tension specimens, and specimens loaded under joint shear. In the second testing phase, five masonry-infilled steel frames were tested with and without retrofit. The composite laminates increased the stiffness and strength and enhanced the post-peak behavior by stabilizing the masonry walls and preventing their out-of-plane spalling. Tests reported in this paper demonstrate the efficiency of composite laminates in improving the deformation capacity of URM, containing the hazardous URM damage, preventing catastrophic failure and maintaining the wall integrity even after significant structural damage.  2005 Elsevier Ltd. All rights reserved. Keywords: Composite masonry; Concrete masonry; Fiber reinforced plastics; Retrofitting; Seismic hazard; Seismic loads; Steel frames

1. Introduction Earthquakes have long been recognized as one of the most damaging natural hazards, along with hurricanes, tornadoes, floods and fire. No other force in nature has the potential to wreak so much havoc in such a short time. Earthquakes typically strike without warning and, after only a few seconds, leave casualties and damage in their wake. Although earthquakes cannot be prevented, the current state-of-the-art in science and engineering provides new tools that can be used to reduce their damaging effects. Through prudent action, the loss of life, serious injury, and property damage as well as social and economic disruptions resulting from earthquakes can be reduced. The principal threat to human life and *

Corresponding author. Tel.: +1 905 525 9140x26109; fax: +1 905 529 9688. E-mail address: [email protected] (W.W. El-Dakhakhni). 0263-8223/$ - see front matter  2005 Elsevier Ltd. All rights reserved. doi:10.1016/j.compstruct.2005.02.017

safety is the shaking damage and the collapse of buildings and other structures that have been inadequately designed or poorly constructed. Major earthquakes can severely disrupt regional or national economic activity by damaging social lifelines such as roads, railways, water, power and communications infrastructures and office and residential buildings. A common type of construction in urban centers for low-rise and mid-rise buildings is unreinforced masonry (URM) walls filling the space bounded by the structural framing members. Although considered non-structural elements, yet under seismic excitation, infill walls tend to interact with the surrounding frame and may result in different undesirable failure modes both to the frame and to the infill wall [1]. In general, URM infill walls have demonstrated poor performance record even in moderate earthquakes. Their behavior is usually brittle with little or no ductility and they, typically, suffer various types of damage ranging from invisible cracking

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459

Nomenclature DH DV c AFRP An As EFRP Es

horizontal extension vertical shortening shear strain cross-sectional area of the FRP per unit length of wall net area of the diagonal tension specimen cross-sectional area of the steel reinforcement per unit length of wall modulus of elasticity of the FRP modulus of elasticity of steel

to crushing and, eventually, disintegration and total collapse. This behavior constitutes a major source of hazard during seismic events and creates a major seismic performance problem facing designers today. Seismic upgrading by adding new structural frames or shear walls, have been proven to be impractical, they have been either too costly or restricted in use to certain types of structures. Other strengthening methods such as grout injection, insertion of reinforcing steel, prestressing, jacketing and different surface treatments were summarized elsewhere [2] and specified by the Federal Emergency Management Agency documents [3,4]. Each of these methods adds considerable mass and stiffness leading to higher seismic loads. They also involve the use of skilled labor and disrupt the normal function of the building. The use of fiber reinforced polymer (FRP) laminates for retrofitting and strengthening is a valid alternative because of their small thickness, high strength-to-weight ratio, high stiffness, and relative ease of application. A strong earthquake introduces severe in-plane and out-of-plane forces to masonry walls which may lead to catastrophic collapse as seen in Fig. 1 during the 1999 Turkey earthquake. However, the majority of work conducted to date [5–11] has been concentrating on the out-of-plane behavior of URM walls strengthened with externally applied FRPs. Infill panels (or large portions of wall) may fall out of the surrounding frame due to inadequate out-of-plane restraint at the frame–infill interface, or due to out-of-plane flexural or shear failure of the infill panel. In undamaged infills, these failures may result from out-of-plane inertial forces, especially for infills at higher story levels and with large slenderness ratios. However, it is more likely for out-of-plane failure to occur after the masonry units become dislodged due to damage from in-plane loading [4]. The work presented herein investigates the effects of applying FRP laminates on the in-plane behavior of URM assemblages subjected to different stress conditions present in masonry infill walls (Fig. 2c). One of the objectives of the present experimental study is to

g h n P Ss t w

vertical gage length diagonal tension specimen height percentage of solid in the masonry unit applied load on the diagonal tension specimens shear stress based on the net area diagonal tension specimen thickness diagonal tension specimen width

investigate the effects of FRP laminates on altering the failure modes and strength and deformation characteristics of different assemblages. Another objective is to demonstrate the potential of the FRP on enhancing the shear and compressive strength of URM infill walls and preventing brittle collapse by means of stabilizing the face shell even after excessive damage. This would also maintain the wallÕs structural integrity and would reduce the possibility of URM walls collapsing and spalling, which, in itself, is a major source of hazard during earthquakes, even if the whole structure remained safe and functioning.

2. Behavior of infill masonry walls Masonry infill walls in frame structures have been long known to affect the strength and stiffness of the infilled-frame structures. In seismic areas, ignoring the frame–wall interaction is not always on the safe side, since, under lateral loads, the infill walls dramatically increase the stiffness by acting as a diagonal strut as seen in Fig. 2a, thus resulting in possible change in the seismic demand due to significant reduction in the natural period of the composite structural system [1]. Also, the composite action of the frame–wall system changes the magnitude and the distribution of straining actions in the frame members, i.e. critical sections in the infilledframe differ from those in the bare frame, which may lead to unconservative or poorly detailed designs. Moreover, these designs may be uneconomical since an important source of structural strength, which is particularly beneficial in regions of low and, sometimes, moderate seismic demand, is wasted. However, URM infill walls exhibit poor seismic performance under moderate and high seismic demand. This behavior is due to a rapid degradation of stiffness, strength and low energy dissipation capacity, resulting from the brittle and sudden damage of the URM infill walls. The rationale behind neglecting infill walls in the design process is partly attributed to incomplete

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Fig. 1. Failure of masonry walls during Turkey earthquake, 1999: (a) out-of-plane failure, (b) in-plane failure and (c) combined in- and out-of-plane failures.

Fig. 2. Behavior of masonry infill walls.

knowledge of the behavior of quasi-brittle materials such as URM and to the lack of conclusive experimental and analytical results to substantiate a reliable design procedure for this type of structures. On the other hand, and because of the large number of interacting parameters, if the infill wall is to be considered in the analysis and design stages, a modeling problem arises because of the many possible failure modes (Fig. 2b) that need to be evaluated with a high degree of uncertainty. This is why it is not surprising that no consensus has emerged leading to a unified approach for the design of infilledframe systems, despite more than five decades of research. It is, however, generally accepted that, under lateral loads, the infill wall acts as a diagonal strut connecting the two loaded corners. However, this is only

applicable to the case of solid infill walls (i.e. with no openings) failing in corner-crushing mode [1].

3. Experimental program The experimental program consisted of two phases. In Phase I, four different types of assemblages (Fig. 2c) were tested under different types of loading conditions representing critical regions in an infill masonry wall. • Axial compression: in-plane, concentric, compressive loads were applied at 90 (normal to the bed joints) and 0 (parallel to the bed joints).

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Table 1 FRP composite and dry fibers properties Composite laminate properties

Dry fibers properties

Ultimate tensile strength in primary fibers direction (MPa) Elongation at break (%) Tensile modulus (GPa) Ultimate tensile strength 90 to primary fibers direction (MPa) Laminate thickness (mm)

309.0 1.6 19.3 309.0 0.25

• Diagonal tension: this is a standard testing procedure used to evaluate the diagonal tensile (or shear) strength of URM and creates a state of stress similar to that occurring in infill walls. • Joint shear: this enabled evaluating the strengthening effect of the FRP laminates against the traditionally weak and brittle horizontal shear slip failure mode. In Phase II, five single-story/single-bay, one-third scale, moment-resisting, structural steel frames infilled with unretrofitted and retrofitted hollow block masonry walls were tested under displacement controlled diagonal loading to evaluate the behavior of the composite frame–wall system. Two series were considered: • Weak frames series: including three identical steel frames tested as a bare (i.e. no infill) frame and two frames were infilled with URM, one of which was retrofitted with FRP. • Strong frames series: including two identical steel frames tested as both were infilled with URM and one of which was retrofitted with FRP.

3.1. FRP selection In order to select an FRP laminate for the URM assemblages, an equivalent-stiffness-based approach [12] was employed. The laminate required was equated to the minimum steel reinforcement ratio of 0.2% (based on the gross cross-sectional area of the wall) according to the requirement of the Masonry Standards Joint Committee [13]. This minimum steel ratio is required in high seismic zones to be distributed between the vertical and horizontal directions in masonry walls. In other words, the required thickness of FRP laminate was determined based on the premise that, the stiffness of the FRP laminate must be at least equal to or greater than the axial stiffness of the reinforcement in the walls. The required thickness of the FRP laminate was therefore calculated by direct scaling of the reinforcement area by the ratio of the elastic moduli of the steel and FRP material as follows, AFRP ¼

Es As EFRP

ð1Þ

Tensile strength (GPa) Tensile modulus (GPa) Ultimate elongation (%) Density (g/cm3) Weight (g/m2)

3.24 72.4 4.5 2.55 295.0

where AFRP is the cross-sectional area of the FRP laminate per unit length of wall, Es is YoungÕs modulus of steel, EFRP is the modulus of elasticity of the FRP laminate, and As is the cross-sectional area of the steel reinforcement per unit length of the wall. The numerical procedure required to select the FRP laminate is given elsewhere [14]. 3.2. Material properties The one-third-scale true-model blocks [15] used in this investigation were replicas of the standard, full-scale 150 mm wide hollow concrete masonry units [16]. The average net-area-based compressive strength of the blocks was 27.87 MPa. The masonry assemblages were constructed using scaled down mortar joints with a nominal thickness of 3.2 mm. To simulate actual construction practice, the mortar mix was designed as Type S mortar [17] and all mortar joints were tooled to a concave profile. The selected FRP was a bi-directional 0/ 90 Glass-FRP with 0.295 kg/m2 of E-glass fibers. The properties of the GFRP composites, given in Table 1, were determined according to ASTM D-3039 specification [18] and were supplied by the manufacturer. However, an average strength of 260 MPa (84% of the specified strength in the GFRP data sheet) with 8.0% C.O.V. was determined by testing five specimens according to the ASTM D-3039 [18]. The steel used for the steel frame sections was of A 36 grade (yield strength 243 MPa) for the Weak, W-Series, frames, and A57250 (yield strength 379 MPa) for the Strong, S-Series, of frames.

4. Phase I testing: assemblages 4.1. Test setup and instrumentation The test specimens were chosen to represent typical loading cases of masonry infill walls as shown in Fig. 2c. A total of 24 1/3-scale specimens were constructed in the laboratory and tested to failure under displacement controlled loading. The overall displacement was measured using Linear Variable Differential Transducers (LVDTs) connected to a PC data acquisition system, which also recorded the applied load on the specimens.

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W.W. El-Dakhakhni et al. / Composite Structures 73 (2006) 458–477 Table 2 Phase I test results Specimen

Results Test number

Strength Individual (MPa)

90U

1 2 3

8.45 8.25 5.94

90R

1 2 3

00U

Average (MPa)

C.O.V. (%)

7.55

18.5

13.10 11.75 11.67

12.17

6.6

1 2 3

5.56 8.12 7.97

7.22

19.9

1 2 3

11.49 11.27 12.51

11.76

5.6

Fig. 3. Assemblages test setup and LVDTs configurations: (a) joint shear and (b) diagonal tension.

00R

Typical LVDTs configurations are shown in Fig. 3 for the diagonal tension and the joint shear specimens.

DTU

1 2 3

0.85 0.79 1.00

0.88

12.5

4.2. Preparation of test specimens

DTR

1 2 3

3.73 4.23 4.07

4.01

6.3

JSU

1 2 3

0.97 0.83 0.66

0.82

18.9

JSR

1 2 3

6.28 7.02 6.30

6.53

6.4

The two axial compression assemblages were of similar dimensions in order to permit direct comparison of their failure loads. Since it was not feasible to cut the 0 assemblages from a built URM walls, the individual blocks for each assemblage were initially cut to shape using a diamond saw. The head mortar joint between the two middle blocks in the joint shear specimens was left unfilled to allow for the specimen to fail in shear. All specimens were constructed with face shell mortar bedding. After air curing for at least 28 days, half of the constructed specimens were retrofitted using two layers of FRP laminates, one on each surface of the specimens. Before applying the FRP laminate, specimen surfaces were first cleaned from mortar protrusions and dust using a wire brush and air blasting, respectively. The epoxy mixture was then applied using a paint roller to both surfaces of the specimen. The pre-cut fabrics were then placed on the wet surfaces and more epoxy was applied to insure complete fabric saturation. The assemblages were tested in accordance with the ASTM E 477-92 [19] and ASTM E 519-81 [20] specifications.

5. Phase I test results The test results are summarized in Table 2, and discussed in the following sections in terms of failure modes, strengths, and deformation characteristics. Each specimen series was assigned a name according to the notation shown with the examples at the bottom of Table 2. The first two characters refer to the axial compression (90 or 00), diagonal tension (DT), or joint

Examples: DTU2 is the second, unretrofitted assemblage tested under diagonal tension; 90R3 is the third, retrofitted assemblage tested under axial compression normal to the bed joint.

shear (JS) assemblages. The third character is assigned one of two letters, either ‘‘U’’ or ‘‘R,’’ indicating whether the assemblage was Unretrofitted or Retrofitted, respectively. 5.1. Axial compression 5.1.1. Failure modes The unretrofitted axial compression assemblages 90U (see Fig. 4a) and 00U (see Fig. 4b) assemblages exhibited typical compression failure modes characterized by vertical splitting along the webs of the two middle blocks [21]. The splitting cracks left the two face shells to deform individually, as shown in Fig. 4c, with a high slenderness ratio. Finally, the specimens totally disintegrated as a result of the out-of-plane buckling and/or spalling of the face shells or a combination of both. Noticeably, all failure modes were brittle and the assemblages disintegrated almost immediately after reaching their respective maximum loads. In contrast, all the retrofitted assemblages exhibited one failure mode initiated by vertical splitting of the

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463

Fig. 4. Failure modes of the unretrofitted axial compression assemblages: (a) Series 90U, (b) Series 00U and (c) web splitting mechanism of face shell mortar bedded masonry [21].

In general, the retrofitted assemblages did not lose all their strength nor disintegrated upon reaching the maximum strength. In fact, in the majority of the tests, a plateau region was attained during which the compressive stress almost stabilized and began to gradually decrease with increased displacement. Such plateau can be regarded as residual strength after failure of the assemblages, an absent feature in the case of URM.

5.1.2. Strength characteristics Table 2 gives the variation of the compressive strengths of the unretrofitted versus retrofitted assemblages. To facilitate comparison, the compressive strengths of the assemblages were calculated as the maximum load-carrying capacity divided by the gross assemblage area perpendicular to the direction of the applied load (6431.0 mm2). It is clear that the values of the coefficients of variation for the retrofitted assemblages are generally lower than those of the unretrofitted specimens (see Table 2). This demonstrates the laminateÕs role in reducing the inevitable variations in URM construction.

Strain (a) 00R3

12.0

Stress (MPa)

interior webs followed by a gradual increase in the load up to the peak load. After reaching the peak load, a sudden bang was heard as a result of the blocks webs completely breaking off the face shells. The specimens continued to carry more load under the displacement controlled loading with a gradual decrease in capacity. All the retrofitted assemblages were reduced to two intact face shells with all interior webs damaged (see Fig. 5).

Stress (MPa)

Fig. 5. Failure modes of the retrofitted axial compression assemblages: (a) Series 90R and (b) Series 00R.

00R1

00R2

10.0 8.0 6.0

00U3

00U2

4.0 2.0 0.0 0.0000

00U1

0.0005

0.0010

0.0015 0.0020 Strain

0.0025

0.0030 0.0035

(b)

Fig. 6. Stress–strain relationships for the axial compression assemblages: (a) Series 90U and 90R, and (b) Series 00U and 00R.

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5.1.3. Deformation characteristics The measured displacement and applied load were used to generate the stress–strain relationships shown in Fig. 6. In general, a good agreement can be observed between the initial slopes for the prisms tested with h = 0 and 90 in both the retrofitted and the unretrofitted series. 5.2. Diagonal tension 5.2.1. Failure modes All the unretrofitted diagonal tension assemblages exhibited shear slip failure along the mortar bed joints. This is attributed to the mortar joint weak bond strength compared to the tensile strength of the concrete blocks. The observed shear slip failure mode was highly brittle and, as soon as shear slip along a bed joint was initiated, the assemblages split into two parts and subsequently

disintegrated. In two out of the three assemblages (DTU1 and DTU2) the shear slip occurred along the middle bed joints (see Fig. 7a), while the third specimen (DTU3) failed along the first bed joint (see Fig. 7b). No signs of cracking or distress were observed prior to failure by shear slip. For the retrofitting assemblages, using the FRP laminates effectively prevented any tension or shear failure modes. All three retrofitted assemblages failed by local crushing of their corners contained within the steel loading shoes. Fig. 7c illustrates the local crushing failures. Due to the compressive stress buildup at the toes, vertical cracking through the webs was observed and extended into the first two courses as shown in Fig. 7d. Other than the local crushing and the minor delamination at the vicinity of the loading shoes, no other signs of distress or cracking were observed in the assemblages.

Fig. 7. Failure modes of the diagonal tension assemblages: (a) specimen DTU2, (b) specimen DTU3, (c) Series DTR corner crushing and (d) Series DTR web splitting.

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5.2.2. Strength characteristics The maximum stresses sustained by the assemblages are given in Table 2. In accordance with ASTM E 519-81 [20] standard specification, the diagonal tensile or shear strength is calculated from, Ss ¼

0:707P An

ð2Þ

where Ss is the shear stress based on the net area, P is the applied load, and An is net area of the specimen calculated as follows:   wþh An ¼ tn ð3Þ 2 where w, h and t are the specimen width, height and thickness respectively, and n is the percentage of solid in the unit, expressed as a decimal. The average strength of the retrofitted diagonal tension specimens was 4.58 times that of the unretrofitted ones and, as expected, the relatively low coefficient of variation is indicative of the role of the laminates in reducing the anisotropy and variability of URM. 5.2.3. Deformation characteristics In accordance with ASTM E 519-81 [20] specification, the shear strain was calculated using the vertical shortening along the compression diagonal and the horizontal extension along the tension diagonal as follows: c¼

DV þ DH g

ð4Þ

where c is the shear strain, DV is the vertical shortening, DH is the horizontal extension, and g is vertical gage. Fig. 8 illustrates the shear stress versus shear strain relationship for the diagonal tension assemblages. Due to the sudden brittle failure mode, obtaining post-peak behavior for the unretrofitted assemblages was not feasible. For the retrofitted specimens, the LVDTs installed at the center of the assemblages at the 102 mm gage length, did not record any appreciable deformations in any of the tests. Furthermore, as the load approached its peak, the long compression LVDT brackets on assemblages

Shear Stress (MPa)

4.5

DTR2

4.0

DTR3

3.5 3.0

465

DTR2 and DTR3 detached due to local delamination and crushing at the vicinity of the bottom loading shoe. The same occurred in assemblage DTR1 towards the end of the test, yet its shear stress versus strain curve showed a load plateau with a slight increase in load. In fact, all the assemblages were able to sustain residual loads under increasing displacement (even though the resulting load plateaus could not be plotted for DTR2 and DTR3). 5.3. Joint shear 5.3.1. Failure modes The unretrofitted specimen failed in a brittle shear slip debonding mode at very low load and displacement levels. This is a result of the weak mortar joint bond strength and the absence of friction resistance due to the lack of compressive stresses normal to the mortar bed joints. The failure was in the form of complete separation in the top and/or bottom mortar joints vicinity (Fig. 9a). This failure mode is highly brittle and occurred without much time elapse between the crack initiation at the block–mortar interface and the consequential debonding of the blocks. In a manner similar to the retrofitted axial compression assemblages, none of the retrofitted assemblages failed by shear slip along the block–mortar interface. At most, minor signs of delamination and stretching of the laminate were observed along a portion of the bed joints. All assemblages ultimately failed after one of the middle blocks cracked through the webs and split open, though remaining attached to the assemblage as shown in Fig. 9b. The splitting of the middle blocks is attributed to the induced lateral tensile stresses developed in the laminates, which resisted the closing of the head joint gap between the top and bottom blocks. The tensile stresses induced in the middle blockÕs face shells eventually resulted in cracking through the webs. This failure might also be attributed to the fact that, with the presence of the laminates and their ability to transfer shear stresses to the middle blocks, the blocks were subjected to a state of stress similar to that occurring in the specimens under compression parallel to the bed joint, and thus the observed failure mode was developed. Nevertheless, this failure mode implies that shear failure can be eliminated and the wall strength would be governed by the compressive strength of the composite prisms.

2.5 DTR1

2.0 1.5

DTU3

1.0 0.5 0.0 0.0000

DTU2 DTU1 0.0005

0.0010

0.0015 0.0020 Shear Strain

0.0025

0.0030

0.0035

Fig. 8. The shear stress versus shear strain relationship for the diagonal tension assemblages.

5.3.2. Strength characteristics The strengths of the three unretrofitted joint shear assemblages are presented in Table 2. In order to determine the joint shear strengths shown in the table, the failure load for each assemblage was divided by the respective net-mortared area. Since face shell mortar bedding was employed, the actual lengths of the

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Fig. 9. Failure modes of the joint shear assemblages: (a) Series JSU and (b) Series JSR.

mortared bed joints less the head joint gap were measured using a caliper and multiplied by the average minimum bottom face shell thickness of 8.7 mm to determine the net joint shear area. The FRP laminates were cut precisely and adhered to cover only the lengths of the mortared bed joints ensuring that the head joint gap between the middle blocks was not obstructed. Therefore, similar to the unretrofitted specimens, the shear area of the retrofitted assemblages is the net-mortared area of the bed joints (excluding the head joint gap area). The increase in the joint shear strength for the retrofitted assemblages was 8.2 times that of the unretrofitted ones.

Fig. 10. The shear stress versus slip relationship for the joint shear assemblages.

5.3.3. Deformation characteristics Fig. 10 illustrates the shear stress versus shear slip behavior of the joint shear assemblages. The unretrofit-

ted assemblages exhibited dramatic load drop after reaching their respective maximum stress at an average

9.0

7.0 6.0 5.0

1.0

Shear Stress (MPa)

Shear Stress (MPa)

8.0

0.6

3.0

JSU3

0.4

2.0

0.2

1.0 0.0 0.00

JSU1

0.8

4.0

0.0 0.000

0.10

0.20

0.30

JSU2 Average Slip (mm) 0.005

0.010

0.40

0.015

0.50

0.020

0.60

0.70

Average Slip (mm)

W.W. El-Dakhakhni et al. / Composite Structures 73 (2006) 458–477

slip of 0.011 mm, with a coefficient of variation of 29.0%. In general, as soon as the retrofitted assemblages reached their respective maximum loads, a sudden load drop occurred which signified the web splitting of one of the middle blocks as discussed above. This was followed by a plateau with a gradually descending strength; the assemblage at this stage was still able to carry more load than the maximum load reached by the unretrofitted specimens. Examination of the failed assemblages revealed that the mortar bond between the top/bottom and middle blocks was damaged and that the laminates were entirely transferring the vertical applied load from the top to the bottom block through the middle blocks. In all the retrofitted assemblages, the laminate was not entirely torn and the assemblage could have resisted further loads. The average slip at the maximum shear stress was 0.379 mm within a coefficient of variation of 17.8%. This is in excess of 34 times that of the unretrofitted joint shear assemblages, thus indicative of the significant deformation capability gained by using the FRP laminate.

467

6. Phase II testing: infilled frames This phase focused on testing of one-third scale, moment-resisting, structural steel frames infilled with unretrofitted and retrofitted hollow block masonry walls. Single-story/single-bay infilled frame subjected to diagonal-compressive loading (as shown in Fig. 11a) were used to evaluate the behavior of the composite frame– wall system. The interaction between the infill wall and the surrounding columns and beams result in unequal contact lengths along the boundaries of the infill with each of the weak and strong frame members. This, in turn, results in different infill contribution with different frames. Table 3 lists the structural properties of the weak and strong frame sections [22] and outlines the five frames tested in Phase II. One-third scaling of the typical clear floor height and column span of the prototype structure was used to obtain the dimensions of the model infilled frame [15]. A clear height between the beams and columns of 1100 mm was used. This is equivalent to a full-scale dimension of 3.3 m. The masonry infill walls were built

AMSLER Loading Jack

Reaction Column W14x90

30.04"

Actuator and Reaction Frame Bracing connected to Top Girder

Applied Load Adjustable Stilt

LC

h

d

SC

L6x6x 3/8 Angle

MT

LT

d

hi gh co ur se s 16

78.49"

1.99 m

L6x6x 3/8 Angle

ST

am Be

0" 00 on 6. cti = e S th = ep D

ST

SC

n um ol C

LC

MC MT LC LT SC ST

MC

e id w

15 6x W

C ol um n

am Be

ks oc bl

W 6x 15

8

Reaction “Strong” Floor

(a)

COMPRESSION DIAGONAL

LT

KEY = Main Compression LVDT = Main Tension LVDT = Long Compression LVDT = Long Tension LVDT = Short Compression LVDT = Short Tension LVDT = Strain gages (adhered to the top of the upper and lower flanges)

T ST SETUP FOR CM & CR FRAMES Drawing Scale 1:20

(b)

LOCATIONS OF STRAIN GAGES AND LVDTs

Fig. 11. Masonry infilled steel frame specimens: (a) test setup, (b) instrumentation.

MT

TENSION DIAGONAL

15 6x

Out-of-plane bracing plates held via C-clamps to the support angles

TENSION DIAGONAL

W

W 6x 15

am Be

C ol um n

MC

am Be

fil In

h= gt en l ill nf 0" =I 0 ht 3.5 g ei 4 lh

C ol um n

n Fr Rea ctio

Load Cell Top Loading Shoe

6. 00 "

=

l

Spherically-seated Actuator Head

l/2

ame Brac ing

COMPRESSION DIAGONAL

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Table 3 Structural properties and test matrix of the steel frames Structural property

Weak (W) frame S3 · 5.7

Strong (S) frame W6 · 15

Area Depth Web thickness Flange width Flange thickness

1077 mm2 76 mm 4 mm 59 mm 7 mm

2858 mm2 152 mm 6 mm 152 mm 7 mm

Strong axis (X-axis) Elastic moment of inertia Elastic section modulus Plastic section modulus

1,049,000 mm4 27,500 mm3 31,950 mm3

12,112,000 mm4 159,300 mm3 176,980 mm3

Weak axis (Y-axis) Elastic moment of inertia Elastic section modulus Plastic section modulus

189,400 mm4 6390 mm3 10,700 mm3

3,879,300 mm4 50,960 mm3 77,840 mm3

Infill type Bare Unretrofitted Retrofitted

Specimen WB WU WR

– SU SR

Steel frame section: ‘‘W’’ weak frame S3 · 5.7 or ‘‘S’’ strong frame W6 · 15. Infill type: ‘‘B’’ bare (no infill), ‘‘U’’ unretrofitted-masonry infill, or ‘‘R’’ retrofitted-masonry infill.

eight blocks wide by 16 courses high. To ensure symmetry in the construction, mortar was packed along all the boundaries between the infill and the confining steel frame with a nominal mortar joint thickness of 3 mm. To identify the different frames tested in Phase II, each specimen was assigned a name according to the notation in the bottom of Table 3. The first character is used to identify the steel section of the bounding frame whether S3 · 5.7 (Weak frame) or W6 · 15 (Strong frame). The second character describes the type of infill, if any, ‘‘B’’ refers to no wall (Bare frame), ‘‘U’’ or ‘‘R,’’ indicating whether the wall was Unretrofitted or Retrofitted, respectively. For the stronger S-frames, the clear height between the beams and the clear span between the columns were similar to those of the W-frames. Moreover, the beam-column connections were also designed as fullmoment-resisting and fabricated using complete-jointpenetration groove welds to weld the beam flanges to the column flanges while 3.0 mm fillet welds were used to weld the beam webs to the column flanges. However, due to the expected high diagonal-compressive loading force, 10 mm thick stiffener plates were welded using 3.0 mm fillet welds between the column flanges in order to prevent premature web buckling at the loaded ends. To maintain symmetry, similar plates were also welded at the four corners of the S-frames.

applied using an AMSLER hydraulic jack with a load capacity of 490 kN and a maximum stroke of 125 mm. A lateral bracing system consisting of four L 6 · 6 · 3/ 8 angles were used to prevent accidental out-of-plane deformations during the in-plane loading of the frames. Fig. 11b illustrates the typical instrumentation installed on the infill wall and bounding frame, as well as the locations of the critical sections along the steel frame where strain gages were placed (an infilled W6 · 15 steel frame is shown in the figure for illustration). All LVDTs, strain gages, and the load cell used to measure the applied compressive load were all connected to a PC data acquisition system. 7. Phase II test results 7.1. WB frame As expected, the frame joints underwent severe rotation and distortion. In addition, both joints along the tension diagonal experienced tearing of the webs as shown in Fig. 12, although no signs of cracking were noted in the welds between the columns and the beams at these moment-resisting connections (thus indicative of the strength of the weld). It was evident from the distorted shape of the frame and the bent columns and beams, that the permanent (i.e., plastic) deformations propagated from the joints inwards. The initial stiffness obtained from the load–deflection curve shown in Fig. 12 was determined to be 2.2 kN/mm. A linear behavior was observed up to a load of a 25.8 kN corresponding to an average top displacement of 11.9 mm. Subsequently, a load plateau of 27.0 kN was attained at a top displacement of 14.0 mm. Attributed to the strain hardening effects, a slight increase in load occurred resulting in an ultimate load of 28.3 kN. The actual load–deflection behavior closely resembles the expected behavior in which an initial linear response occurs until initiation of yielding followed by a plateau then a slight load increase to attain the ultimate failure load. 7.2. WU frame The maximum load-carrying capacity attained by the infilled frame was 104.0 kN. A stepped diagonal crack

6.1. Test setup and instrumentation The general test setup and loading assembly are shown in Fig. 11a. The compressive top load was

Fig. 12. Load–deflection relationship for frame WB.

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Load (kN)

100 WU

80 60 40

WB

20 Applied Load

0 0.0

20.0

40.0 Deflection (mm)

60.0

80.0

Fig. 13. Load–deflection relationship for frames WU and WB.

was observed at the center of the infill panel along the compression diagonal as shown in Fig. 13, was announced by an audible bang and occurred at a load of 7.8 kN corresponding to 7.5% of the maximum attained load. The load–deflection relation obtained for frame WU is shown in Fig. 13. The load versus deflection curve of the bare frame tested earlier is reproduced on the same plot for comparison. Initially, the load–deflection curve was characterized by a steady rise as shown in Fig. 13. Prior to attaining the first peak at 97.7 kN at a corresponding deflection of 1.9 mm and shortly afterwards, cracking noises were continuously heard although no visible cracks were observed, thus indicating possible damage in the interior webs of the masonry infill panel resulting in a series of quick load descents and ascents. Shortly afterwards, a major off-diagonal crack parallel to the initial toothed crack at the center of the infill panel was observed. This crack occurred at a trough in the load–deflection curve at approximately 5.1 mm. Unlike the initial toothed crack, this crack propagated vertically through the

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blocks and mortar joints. As the infill panel further readjusted indicating redistribution of the transferred load, the load steadily rose to reach the maximum load carried by this infilled frame in spite of the occurrence of local cracking due to crushing of the infill in the vicinity of the bottom loading shoe at approximately 7.6 mm. At 12.7 mm, crushing and cracking was observed in the infill panel near the top loading shoe. Subsequently, a steady decline in the load-carrying capacity of the frame occurred as more off-diagonal cracks started appearing on both sides of the first toothed crack in addition to spalling of the block face shells near these cracks. At approximately 15.0 mm, cracks along the infill bed joints were observed and propagated at mid-length of the frame columns until a significant portion of the infill panel face shell near the mid-length of the left frame column separated (approximately at 28.0 mm) as shown in Fig. 14a. Face shell spalling continued rapidly (at 33.0 mm) resulting in severe deterioration and cracking in the infill panel until, finally (at 56.0 mm), was attained signifying the beginning of a load plateau. In turn, this marks the point after which the infill panel ceased to contribute in resisting the applied load. Extensive local crushing and cracking at the toe of the infill panel near the bottom loading shoe is shown in Fig. 14b. The severity of the damage at the center of the masonry infill is illustrated in Fig. 14c. From 56.0 mm onwards, the steel frame was entirely carrying the applied load and was deformed severely as shown in Fig. 14d. Similar to the WB frame, the joints of the steel frame underwent severe plastic rotation at the joints along the tension diagonal,

Fig. 14. Damage of frame WU: (a) at the left column, (b) at the wall toe, (c) at the infill center and (d) at the beam-column joint.

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in addition to local web buckling in the columns and beams at the loaded joints. The infilled WU frame attained a maximum load-carrying capacity of 104.0 kN which represents an increase of 267.5% compared to the capacity of the bare WB frame. The initial stiffness of the WU frame measured as the secant stiffness at 50% of the maximum load-carrying capacity was determined as 55.7 kN/mm which is 25 times that of the bare frame. The significant increase in WU frameÕs load-carrying capacity and initial stiffness compared to the WB frame was expected, particularly in view of the fact that the infill panel is relatively stiff compared to the frame. Ultimately, upon attaining a load plateau at a displacement of 57.4 mm signifying the end of the infillÕs participation in in-plane load resistance, the plastic load-carrying capacity of the WU frame was determined from the test as 28.2 kN. This is comparable to the bare frameÕs ultimate load capacity obtained previously from the WB frame test. 7.3. WR frame The load versus deflection curves for the WR frame is shown in Fig. 15a. The maximum load attained by the frame was 218.9 kN which represents increases of 7.7 times and 2.2 times the maximum loads attained by the WB and WU frames respectively. At an applied load of 182.4 kN corresponding to 83.3% of the maximum

load-carrying capacity, some hairline cracks were observed in the blocks near the vicinity of the top loading shoe. These cracks were visible underneath the clear laminate adhered on the exterior of the masonry infill panel. As the load began to decrease, cracking noises and clicks were heard until suddenly at a load of 175.1 kN, corresponding to 80.0% of the ultimate load on the descending branch of the load curve, the interior webs near the top portion of the infill panel were damaged causing the separated retrofitted face shells near the top loading shoe to snap outwards and moved outside the flanges of the frame members (Fig. 16a). Unfortunately, shortly before and after this outward ‘‘burst,’’ the buckled face shell brushed against the main and infill compression LVDTs, thus preventing further recording of displacement. Minor signs of delamination along the second bed joint on the backside of the infill panel were evident. There were no signs of distress in the steel frame, clearly indicating that the applied load was primarily endured by the retrofitted infill panel with minimal contribution from the steel frame. A thorough understanding of the behavior and response of the retrofitted infill panel was further facilitated upon its removal from the bounding steel frame thereby enabling a closer inspection. Fig. 16b shows the wall separation in the left side of the frame. The retrofit technique using FRP laminates was very successful

Fig. 15. Frame WR: (a) load–deflection relationship, (b) delamination zone and (c) damaged webs region.

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in preserving the integrity of the highly brittle masonry. The fact that the panel, simulating a story-high wall, was removed in one intact piece (in spite of some damage to the interior webs) is testimony to the beneficial effect of retrofit with FRP overlay. At the toes of the infill panel within the vicinity of the loaded corners of the frame, all interior webs were damaged. As shown in Figs. 15c and 16c, the web damage extended inwards towards the center of the panel to approximately one-quarter of the diagonal length. Web damage was also evident along the perimeter of the panel which was in contact with the frame members. The web damage was minimal near the corners of the tension diagonal. However, the separated face shells were held intact by the strong laminate and, in general, there was minimal (if any) delamination between the overlay and the block face shells (except in the location shown in Fig. 15b). No web damage was evident at the center of the infill. Furthermore, unlike in the unretrofitted-masonry infill wall of the WU frame, the laminate successfully prevented the occurrence of any diagonal

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tension cracks or shear slip along the bed joints in the infill. As shown in the various assemblage tests discussed in Phase I of the experimental program, the FRP laminate basically suppressed any tension and shear failure modes in the masonry by reinforcing the weak mortar joints. Thus, the resulting retrofittedmasonry infill wall has been transformed into two very strong face shells connected by masonry webs which are considered as the weak elements in the assembly. The secant stiffness at 50% of the maximum load, the initial stiffness of the BR frame is 131.4 kN/mm which represents increases of 58.7 times and 2.4 times the initial stiffness values of the WB and WU frames respectively. The peak load was reached at a compressive deflection of 5.6 mm. Similar to the WU frame, the recorded deflections along the compression diagonal in the direction of the applied load were greater than those along the tension diagonal. This is attributed to the local cracking at the infillÕs loaded toes resulting in a reduced stiffness along the loading direction.

Fig. 16. Damage of frame WR: (a) out-of-plane wall burst, (b) infill wall separation and (c) extent of web splitting.

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Ultimately, as the infill panel was no longer in any effective contact with the bounding steel frame, a load plateau was attained which represents the bare frameÕs plastic load capacity. The load stabilized at a value of 28.9 kN which is comparable to the WB frame capacity. 7.4. SU frame Plot of the applied top load versus deflections for the SU frame is shown in Fig. 17. The initial secant stiffness of the SU frame was 91.4 kN/mm. The ultimate loadcarrying capacity of the SU frame determined in the second test was 284.4 kN. The unretrofitted-masonry infill panel remained crack-free up to an applied diagonal load of 122.4 kN corresponding to 43.0% of the ultimate load-carrying capacity of the SU frame. Thereafter, a longitudinal crack at the middle of the panel occurred similar to the WU frame. However, unlike the toothed crack in the WU frame, which propagated along the head and bed joints around the masonry units, the middle crack in the SU frameÕs panel extended through both the units and the mortar joints. As loading progressed, the middle crack extended further in addition to the formation of some off-diagonal hairline cracks and a short bed joint crack above the first masonry course. Near the peak load, signs of crushing of the boundary mortar joint between the steel frame and the infill panel in the vicinity of the loaded corners were observed. Moreover, a hairline separation crack between the panel and the frame at the tension corners was observed to extend approximately three courses long. The cracking pattern of the masonry infill wall resembles that encountered in the WU frame test in which a central crack is first initiated along the loaded diagonal of the wall followed by the formation of some off-diagonal cracks. In the second test attempt, the existing but closed hairline cracks resulting from the first test widened as the frame reached a first peak load at 266.9 kN at a corresponding deflection of 6.8 mm. Shortly before reaching the peak load, minor cracking was observed in the infillÕs toe near the bottom loading shoe. As the frame was pushed further in spite of the decreased load resistance, small off-diagonal cracks began forming on the left and right sides of the central crack. These cracks assisted in the redistribution of the load

Load (kN)

300.0 250.0 200.0 150.0 100.0 50.0 0.0 0.00

20.00

40.00

60.00

80.00

Deflection (mm)

Fig. 17. Frame SU load–deflection relationship and diagonal cracking.

within the infill panel as it adjusted to bear against the deforming shape of the steel frame. Suddenly, face shell spalling occurred in the tension corner regions (to the left and right of the central crack where extensive offdiagonal cracks were occurring) as shown in Fig. 18a. This occurred at an approximate SU deflection of 9.0 mm. In fact, this served as a major indicator of the shift in load resistance between the infill panel and the steel frame. As the frame was still being pushed diagonally, due to incompatible deformations between the steel frame and the masonry wall, the infill wall was quickly losing structural integrity accompanied with rapid face shell spalling and collapse of massive ‘‘chunks’’ of the upper region of the wall (Fig. 18b). At the end, only the lower three courses of the masonry wall remained standing on the lower beam and column of the frame as shown in Fig. 18c. As the load increased, the SU frame was simply behaving as a bare W6 · 15 moment-resisting frame and yielding commenced at the beam-column joints. The formation of plastic hinges, eventually leading to a plastic collapse mechanism, characterizes the failure mode of the CM frame in which the masonry infill panel did not increase its load-carrying capacity. However, failure of the masonry infill panel is attributed to a combination of toe crushing (characterized by local compressive crushing of the masonry at the vicinity of the loaded corners) and diagonal-compression failure at the center of the panel (characterized by the formation of extensive diagonal and off-diagonal longitudinal cracks). The frame was unloaded after the full stroke of the loading actuator was consumed. 7.5. SR frame Fig. 19a shows the load–deflection relationship of the SR frame which had a maximum load capacity of 343.0 kN and an initial stiffness of 262.7 kN/mm. At 0.8 mm, an audible bang was heard although no crack was detected visually. Characterized by a small shift in the load–deflection curve, the bang suggested the occurrence of a crack in the interior webs, possibly at one of the loaded toes of the infill wall. At 2.3 mm, the crack location most probably occurred in the top toe. At a deflection of 1.98 mm, the load increased further until it reached the first peak of 270.1 kN. At this point, a greater load bang was heard and crushing at the top loaded toe of the infill panel was observed. As in the retrofitted assemblages, crushing at the top toe of the retrofitted-masonry infill panel involved damage of the interior webs leading to the laminated-face shells snapping outwards. The damage extended along the infill-frame boundary for a length of approximately six courses (three block-lengths) and only one course wide (one half a block-length) as shown in Fig. 19b. A consequential loss in load capacity occurred but gradually

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473

Fig. 18. Damage of frame SU: (a) face shell spalling in the diagonally cracked region, (b) collapse of the upper infill region and (c) infill wall remnants.

Fig. 19. Frame SR: (a) load–deflection relationship, (b) damage progress and (c) final damaged zones.

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increased as the frame was further loaded. Similar crushing also occurred at the bottom loaded toe of the infill panel (Fig. 19b). Separation between the infill panel and the steel frame at the tension diagonal corners occurred as loading progressed. At approximately 22.0 mm, the separation gap was clearly visible as shown in Fig. 20a. As the frame was pushed further beyond 22.0 mm, the separation between the frame and the infill increased. The extent of toe crushing, which is defined as splitting of the face shell and at times accompanied by minor delamination between the overlay and the face shell itself, also increased (Fig. 20b and c). The steel frame was considerably deformed with significant plastic rotation at the tension joints as shown in Fig. 20d. Examination of the infill panel at the end of the test (Fig. 20e) indicated

that, other than minor delamination at the infill-frame boundaries and toe crushing and in spite of the separation between the infill and the frame at the tension diagonal corners, the central region of the wall was intact without any cracking or damage. A schematic diagram illustrating the state of the retrofitted infill wall at the end of the test is shown in Fig. 19c. Even though the load–deflection curves indicated that a load plateau was reached, it was decided to further load the SR frame. This decision was triggered by the fact that the plateau occurred at load of 339.0 kN which is 21.0% greater than the expected plastic load-carrying capacity of the bare W6 · 15 frame of 280.4 kN as determined experimentally from the prior SU frame test, thus suggesting that the retrofitted infill wall still contributed to load resistance.

Fig. 20. Frame SR damage: (a) at left side, (b) at top loaded corner, (c) at bottom loaded corner, (d) at beam-column joint and (e) final damaged configuration.

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8. Summary of Phase II test results Beside local toe crushing, secondary signs of distress resulting from severe face shell splitting such as tearing of the laminate or minor delamination between the block face shells and the laminates were the only observed damages as the FRP retrofitted infilled-frames were pushed to severely deformed configuration. The frames with the retrofitted infill walls depicted similar behavior throughout the entire loading history. In both the WR and SR tests, as soon as local crushing occurred at the wallÕs corners, clearly visible and wide separation gaps between the panel and the frame constantly increased unlike in the WU and SU frames. Unlike the unretrofitted-masonry infill walls which disintegrated soon after the infilled-frame system reached its peak load, the retrofitted infill walls remained supported within the bounding steel frame until the end of the loading and even after attainment of load plateau which signaled that the frames reached the plastic load-carrying capacity of the bare steel frame. This behavior demonstrates the superior contribution of FRP laminates in altering the brittle hazardous behavior of URM infill walls to a ductile and damage-tolerant wall with apparent postpeak capacity and energy dissipation capabilities. Table 4 summarizes the maximum diagonal-compressive load sustained by the five tested frames. In addition, the plastic load capacity of the bare W6 · 15 which was experimentally determined through testing the SU frame after the remains of the infill panel were removed (the third test of the SU frame), is also included in the table. The percentage and the corresponding multiple increases in the load-carrying capacity compared to that of the bare frame and the unretrofitted-masonry infilled frame for each of the two steel frame types are also calculated and presented in Table 4. Table 4 also compares the initial secant stiffness of the five tested frames in this study. The stiffness values were calculated as the slope of line joining the origin and the

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point at 50% of the ultimate load using the applied diagonal load versus in-line compressive displacement curves. Although a bare W6 · 15 steel frame was not tested as a separate frame, its stiffness which is shown in Table 4 was experimentally determined from the second test of the SR frame whose infill panel was separated from the frame along the majority of its perimeter and sustained local damage at its loaded toes whereas the steel frame did not experience any distress in the prior test. For each frame within the two main steel frame types, the increases in stiffness compared to the bare frame and the first test of the unretrofitted-masonry infilled frame are computed and presented in the table.

9. Conclusions This paper presents an experimental investigation on the retrofitting of concrete masonry infill walls using FRP laminates, which provides a strengthening alternative for URM infill walls. The relative ease with which FRP laminates can be installed on the walls makes this form of strengthening attractive to the owner, considering both reduced installation cost and down time of the occupied structure. Another reason is to comply with new seismic codes requirements without the need to demolish the whole wall and rebuild it. The following conclusions resulted from Phase I of the investigation: 1. The laminates significantly increased the load-carrying capacity of the masonry assemblages exhibiting shear failures along the mortar joints (joint shear and diagonal tension). The average joint shear strength of the retrofitted specimens was equal to eight times that of their unretrofitted counterparts. 2. The unretrofitted axial compression assemblages failed suddenly and disintegrated totally upon reaching peak stress. However, the FRP supplied the tensile strength required to stabilize the out-of-plane

Table 4 Phase II test results Frame

W-Frames (S3 · 5.7) S-Frames (W6 · 15) a

Maximum load (kN)

WB WU WR SB SU SR

28.3 104.0 219.0 280.4b 284.4 343.1

% Increase compared to

XÕs increase compared to

Bare frame

Bare frame

267.5% 673.9% 1.4% 22.4%

Unretrofitted infilled frame

Unretrofitted infilled frame

110.6%

3.67 7.74

2.11

20.6%

1.01 1.22

1.21

Initial secant stiffnessa (kN/mm)

2.2 55.7 131.4 20.9c 91.4 175.1

XÕs Increase Compared to Bare frame

Unretrofittedmasonry infilled frame

24.92 58.74

2.36

4.38 8.40

1.92

Initial secant stiffness determined from the load–deflection curve of the tested frames as the slope of the line joining the origin and the point at 50% of the maximum load-capacity. b Plastic load-carrying capacity of the bare W6 · 15 frame was experimentally determined from the third test of the SU frame after the remains of the infill wall were removed entirely. c Initial secant stiffness of the bare W6 · 15 frame was experimentally determined from retesting the SR frame.

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buckling of the individual face shells, thus preventing brittle failure after webs splitting and allowing the specimen to carry more loads. This prevented catastrophic failure of the masonry–FRP composite assemblages compared to their URM counterparts. 3. The FRP laminates resulted in a gradual prolonged failure, and a stable wall with noticeable structurally integrity and residual strength even after failure. Thus, the long known hazard problem associated with URM can be eliminated using the proposed retrofit technique. In seismic zones, the prevention of the brittle failure mode is highly desirable since it provides a means for energy dissipation and consequently reduces the seismic forces on the frame structure. The following conclusions resulted from Phase II of the investigation: 4. Retrofitting the infill panel with externally, epoxybonded FRP laminates resulted in an increase in load-carrying capacity of 2.1 and 1.2 times that of the corresponding unretrofitted-masonry infilled frames for the W-frames and the S-frames respectively. 5. Even in the S-frames whose load capacity was not significantly increased due to retrofit of the infill panel, the laminates were able to completely alter the deformation characteristics and behavior of the wall itself. In the unretrofitted-masonry infilled frames, the walls were completely destroyed and the blocks fell out-ofplane which in real life poses a hazard to buildingsÕ occupants. The failure mode of the two unretrofitted frames was characterized as corner-crushing and diagonal-compression respectively. In the retrofitted-masonry infilled frames, no signs of diagonal cracking were observed and both frame types failed due to local crushing at the loaded corners. Examination of the retrofitted panel indicated that the central region remained intact and that the majority of the damage occurred at the outermost perimeter and at the loaded corners where the inner webs of the blocks cracked resulting in the formation of separate laminated-face shells. 6. The retrofitting technique maintained the walls structural integrity and prevented collapse and debris fallout. The FRP laminates contained and localized the damage of the URM walls even after ultimate failure and no signs of distress were evident throughout the wall except at the vicinity of the corners and around the openings. In contrast to the URM walls, the strengthened walls were stable after failure. In a real building, this can reduce the seismic hazard associated with the wall tipping off or falling out of the frame, and eliminate injuries or loss of lives and properties due to the wall collapse. This would also

maintain the wallÕs structural integrity and would reduce the possibility of URM walls collapsing and spalling, which, in itself, is a major source of hazard during earthquakes, even if the whole structure remained safe and functioning. 7. The masonry–FRP composite walls do not fail catastrophically as their URM counterparts. The FRP laminates resulted in a gradual prolonged failure and a stronger wall with more energy dissipation and apparent post-peak strength. This should result in a higher response modification factor than that typically selected for the analysis of URM structures. 8. By supplying the shear strength at the mortar joints, FRP laminates can serve as external reinforcement for unreinforced or under-reinforced masonry walls, thus providing a quick and cost-effective solution to conform to the more restrict emerging seismic codes requirements.

Acknowledgments The work presented herein was supported under Grant No. CMS-9730646 from the National Science Foundation (NSF). The results, opinions, and conclusions expressed in this paper are solely those of the authors and do not necessarily reflect those of the NSF. The authors would like to gratefully acknowledge assistance of Edward Fyfe, Peter Milligan and Sarah Cruickshank, Fyfe Co. LLC, California, for providing the FRP, and John Sabia, D.M. Sabia Co., Pennsylvania for providing the mason. The authors would also like to thank Mr. Omar El-Dakhakhni for his assistance.

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