Out-of-plane Proof Testing of Masonry Infill Walls

Accepted Manuscript Out-of-plane Proof Testing of Masonry Infill Walls

Dmytro Dizhur, Kevin Walsh, Ivan Giongo, Hossein Derakhshan, Jason Ingham PII: DOI: Reference:

S2352-0124(18)30067-5 doi:10.1016/j.istruc.2018.07.003 ISTRUC 297

To appear in:

Structures

Received date: Revised date: Accepted date:

18 November 2017 2 June 2018 12 July 2018

Please cite this article as: Dmytro Dizhur, Kevin Walsh, Ivan Giongo, Hossein Derakhshan, Jason Ingham , Out-of-plane Proof Testing of Masonry Infill Walls. Istruc (2018), doi:10.1016/j.istruc.2018.07.003

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ACCEPTED MANUSCRIPT OUT-OF-PLANE PROOF TESTING OF MASONRY INFILL WALLS Dmytro Dizhur1, Kevin Walsh2, Ivan Giongo3, Hossein Derakhshan4, and Jason Ingham1

ABSTRACT

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Proof testing of multiple fired-clay-brick unreinforced masonry (URM) infill walls set within

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reinforced concrete frames was undertaken using airbags to simulate out-of-plane (OOP) loading. The proof testing was conducted to provide engineers in various research and practitioner roles with

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verified wall behaviour for the purpose of seismic assessment of buildings. A total of 19 tests were

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performed in six buildings. It was observed that two-way OOP flexure can substantially improve the OOP load-carrying capacity of tested infill walls compared to one-way vertical OOP flexure and that

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boundary restraints and presumed ‘arching’ action from the building frame can significantly increase the OOP capacity of URM walls. In addition, the effects of simulated in-plane damage on the OOP

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capacity of a URM infill wall were investigated, and it was found that the damage reduced the OOP

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strength by up to 40%. On-site proof testing is demonstrated as a simple and cost-effective way to establish actual wall lateral capacities in cases where boundary conditions cannot be clearly established

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and existing analytical models predict low lateral capacities.

Department of Civil and Environmental Engineering, University of Auckland, New Zealand, [email protected], [email protected] Department of Civil & Environmental Engineering & Earth Sciences, University of Notre Dame, Indiana, United States, [email protected]; Frost

Engineering and Consulting, Mishawaka, Indiana , United States, [email protected] 3 4

Department of Civil, Environmental and Mechanical Engineering, University of Trento, Italy, [email protected], School of Civil, Environmental and Mining Engineering, University of Adelaide, Australia, [email protected]

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INTRODUCTION

Out-of-plane (OOP) failure of loadbearing unreinforced masonry (URM) walls often occurs during moderate and severe earthquake shaking (Brodie and Harris 1933; Shepherd et al. 1990; Somers et al. 1996; Verderame et al. 2009; Moon et al. 2012; Giaretton et al. 2016), and in structural engineering

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assessments such walls are routinely identify as amongst the elements that are most vulnerable to OOP lateral loading generated by earthquakes, wind, or blasts. In contrast to loadbearing URM walls,

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non-loadbearing masonry infill walls typically perform well when subjected to OOP loads [see Figure

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1], with comparatively few collapse cases despite often being identified as vulnerable by engineering

(b) Single-storey building in Christchurch, New Zealand with no damage to masonry infill (March 2011)

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(a) Four-storey building in Christchurch, New Zealand with negligible damage to masonry infill (March 2011)

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analyses.

(c) Four-storey building in Kathmandu, Nepal with no damage to masonry infill (May 2015)

Figure 1. Examples of good structural performance of URM infill walls following large magnitude

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earthquakes

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Following the 2009 L’Aquila earthquake in Italy, it was observed that some masonry infill walls collapsed primarily due to OOP mechanisms, resulting from inadequate or absent cavity ties between the inner and outer wythes of masonry (Verderame et al. 2009; Braga et al. 2011) [see Figure 2 (a)]. Further, following the 2010–2011 Canterbury earthquakes in New Zealand and the 2015 Gorkha earthquake in Nepal (Galetzka et al., 2015, Avouac et al. 2015, Dizhur et al. 2016), a small number of masonry infill walls were observed to have collapsed due to OOP mechanisms [see Figure 2(b),(c)]. Investigators compiling reconnaissance reports following the 1989 Loma Prieta (Shepherd et al. 1990) 2

ACCEPTED MANUSCRIPT and the 1994 Northridge (Somers et al. 1996) earthquakes made similar observations regarding the

(b) Collapse of masonry infill walls in Christchurch, New Zealand (March 2011)

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(a) Collapse of masonry infill walls in L’Aquila, Italy (Credit: University of Padova, 2009)

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potential for slender brick masonry infill walls to collapse if not properly restrained.

(c) OOP collapse of masonry infill walls in Kathmandu, Nepal (May 2015)

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Figure 2. Examples of OOP damage and collapse of masonry infill walls during earthquakes In response to the observations described above, a simple and cost-effective in-situ proof testing

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experimental program was undertaken to establish the OOP capacity of non-loadbearing masonry infill walls and to provide detailed information to guide structural engineering practitioners assessing the

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OOP behaviour of this wall type. A total of 19 tests on masonry walls were performed in six buildings

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utilising an approach wherein lateral forces were applied using a system of airbags to simulate distributed OOP forces. This approach is consistent with the testing procedures recommended by the

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American Society of Civil Engineers (ASCE 2014) and previously utilised by Abrams et al. (1996) and Derakhshan et al. (2014a, 2014b). The test samples represent a variety of geometries, boundary

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conditions, pre-test damage states, and material properties. The objectives of the OOP airbag proof testing reported herein were to provide a basis of knowledge regarding: 

The execution of and benefits to engineers and building owners of proof testing masonry infill/partition walls as part of a larger detailed seismic assessment programme.

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The OOP behaviour of masonry infill/partition walls in one-way vertical flexure as well as two-way flexure and the effects of boundary restraint conditions on the OOP behaviour of masonry infill/partition walls.

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ACCEPTED MANUSCRIPT 

Comparisons of experimentally attained capacities to anticipated seismic demands in regions of moderate to high seismicity.

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The effects of in-plane shear damage on OOP initial stiffness and ultimate strength and drift capacities of masonry infill/partition walls.

Experimental results attained from proof tests may lead to more efficient retrofit design of building

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components and subsequent reductions in cost, time, energy, solid waste production, and other

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are relatively costly and invasive (e.g., steel-framed strongbacks).

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resources associated with deconstruction, partial reconstruction/replacement, or retrofit solutions that

TESTED WALLS

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This section provides a brief description of each tested building and the associated test walls. The

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Weir House (WH) estate in Wellington, New Zealand was constructed in 1932 [see Figure 3(a)]. Internal masonry partition walls are constructed of 90 mm thick terracotta masonry lined with cement plaster on both sides [see Figure 3(b)]. One test wall (WH1) is a partition wall located on second

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storey, and the other two test walls (WH2 and WH3) are partition walls located at the top storey of the building. Partition wall WH1 is restrained by a reinforced concrete (RC) column on one side (vertical edge), a masonry return wall on the other side, and RC floor slabs above and below. WH2 is restrained

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by an RC column on one side (500 mm x 700 mm, perpendicular to the plane of the wall), a URM

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return wall on the other side, an RC floor slab below, and only drywall (i.e., plasterboard, wallboard, or gypsum board) at the top (i.e., the top edge of the wall was effectively unrestrained). WH3 is restrained by an RC shear wall on one side (vertical edge), a timber wardrobe on the other side, an RC floor slab below, and only drywall at the top (i.e., as with WH2).

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(a) Front view (b) Wall cross-section Figure 3. Weir House (WH)

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The Wellington Oriental Bay (WO) apartment building was constructed in the early 1900s as a two-storey building with predominantly loadbearing URM perimeter walls [see Figure 4(a)]. An RC

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ring beam extends around the perimeter of the building at the level of the second-storey floor

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diaphragm and roof. Both test walls are located on the first storey. Wall WO1 is restrained by URM return walls on both sides (vertical edges), timber framing above (lateral restraint only,

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non-loadbearing), and an RC/URM foundation system below. Wall WO2 is restrained by a URM return wall on one side [the return wall is test wall WO1, see Figure 4(b)], a door frame on the other

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side, loadbearing timber framing above (applying some overburden load to the wall), and an RC/URM

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foundation system below.

(a) Front and side view

(b) Wall WO2 (wall WO1 is the return wall to the left of WO2) Figure 4. Wellington Oriental Bay apartment building (WO)

A third building, the Wellington Railway Station (WR), was officially opened in 1937. The main structure has five and six storeys and concrete-encased steel moment-resisting frame construction [see Figure 5(a)] with RC floor slabs atop RC concrete pile foundations with exterior partially reinforced masonry cladding and interior URM partition walls. All test walls in this building rest on RC floor 5

ACCEPTED MANUSCRIPT slabs, and most (WR1, WR2A, WR2B, WR3, WR4) are bound by 127 mm thick RC floor slabs above. The remaining test walls (WR5 and WR6) are restrained laterally at the top by timber roof framing, which does not impose any overburden load onto the walls. Most of the test walls (WR1, WR2B, WR3, WR4, WR6) are unrestrained on the side (vertical) edges with either saw cuts made prior to testing or tall door openings. The remaining test walls (WR2A and WR5) are restrained on the side

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edges by a URM return wall, a URM pier, or a short door opening [see Figure 5(b)–(f)].

(c) WR3

(d) WR4

(f) WR6 (WR5 similar)

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(b) WR1 (WR2B similar)

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(a) Front view

Figure 5. Wellington Railway Station (WR)

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A fourth building, the automotive garage (AG), was located on Victoria Street West in the Auckland Central Business District prior to its demolition. The tested components of the building were originally constructed in 1958. The main building structure consisted of a tall, one-storey, concrete-encased steel moment-resisting frame. On the exterior walls, two single wythes of clay hollow-core brick separated by a cavity served as infill within the frame bays [see Figure 6(a)]. The exterior wythe was removed from both test panels (AG1 and AG2) prior to testing the interior brick wythe [see Figure 6(b)]. Both test panels rested on a thick RC slab on grade and were bound by concrete-encased steel columns on 6

ACCEPTED MANUSCRIPT both side (vertical) edges (265 mm x 305 mm, perpendicular to the plane of the wall) and a shallow RC beam above (150 mm x 280 mm horizontal/perpendicular to the plane of the wall). The bounding

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columns had small concrete shear keys plus steel wires extruding into the mortar of the masonry walls.

(b) AG1 (left) and AG2 (right)

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(a) Front view

Figure 6. Automotive garage building (AG)

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A fifth building, a retail building (AO) located in Orakei, Auckland, was originally constructed in

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1938 [see Figure 7(a)]. The main structural system of the building consists of regularly spaced RC frames, exterior URM cavity infill walls, and interior URM partition walls. The floor diaphragms

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consist of RC slabs at the ground and first floors and a suspended timber floor at the basement level.

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Interior partition test wall AO1 is located in the basement level and hence rests on the suspended timber floor. Test wall AO1 is bounded on the side (vertical) edges by RC columns (350 mm x 350 mm on the interior and 300 mm x 300 mm on the exterior), with an RC beam above (375 mm x 300 mm,

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horizontal/perpendicular to the wall plane [see Figure 7(b)]).

(a) Rear view

(b) AO1

Figure 7. Retail building in Orakei, Auckland (AO) A sixth building (AK) is located on Kingston Street, Auckland [see Figure 8(a)] and was originally 7

ACCEPTED MANUSCRIPT constructed in 1927 as office and warehouse space. The four-storey RC frame building is part of a building row along Kingston Street. Test walls AK1 and AK2 consisted of two 75 mm thick solid clay brick wall layers separated by a cavity with metal wire ties interconnecting the cavity [see Figure 8(b)]. All four sides of test wall AK1 are restrained by RC members, and the wall was tested in the as-built condition with original ties. The top and bottom edge restraints of AK1 and AK2 consist of

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300 mm x 475 mm RC beams. Test wall AK2 was vertically cut on two sides to form a 1400-mm-wide

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wall panel, and all connecting wall ties were removed. The top and bottom boundary conditions remain

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in their original condition.

(b) AK1 (right) and AK2 (left)

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(a) Front view

Figure 8. Kingston Street building, Auckland (AK)

TESTING DETAILS

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The experimental proof testing programme consisted of 19 tests performed on 16 walls. A summary of geometric details and boundary conditions of the test wall specimens is given in Table 1. Note that walls with test identifications (IDs) ending with a letter (e.g., A, B, or C) were tested multiple times with different levels of simulated damage or changes in boundary conditions. As noted in Table 1, test walls WO1B, WO1C, and AG2 were prepared with simulated damage by saw cutting 50-mm-deep ‘cracks’ into the walls’ compression sides prior to testing [see Figure 9(a)-(c)]. Test wall AO1 was saw cut 50 mm deep through the bottom masonry course to simulate the effects of the smooth damp-proof 8

ACCEPTED MANUSCRIPT lead course on the exterior wythes of the perimeter walls. Most of the test walls in the WR building (WR1, WR2B, WR3, WR4, WR6) were tested unrestrained on the side (vertical) edges utilising either saw cuts made prior to testing or tall door openings [see Figure 9(d)] to conservatively simulate one-way vertical flexure for purposes of analysis elsewhere in the building. Plaster was retained on both sides of the test walls in the WH and WO buildings and on various other walls in this testing

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programme but otherwise ignored in analysis (based on the assumption that the plaster’s marginal

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contribution to OOP strength would be approximately equally offset by its marginal self-weight).

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Hence, for purposes of estimating the self-weight of the test walls, only the brick thicknesses listed in

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Table 1 were considered, as all test walls in this programme consisted of a single brick wythe.

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ACCEPTED MANUSCRIPT Table 1. Summary of test wall geometries, boundary conditions, and preparations Test ID

Length (mm)

Wall height (mm)

Brick thickness (mm)

Top edge restraint

WH1

4100

3600

95

RC slab

WH2

3850

2730

95

Gypsum board (free)

WH3

3480

2730

95

Gypsum board (free)

WO1A

3900

2740

110

Timber (lateral only)

Side (vertical) edge restraints

Bottom edge restraint

RC column and URM return wallURM return RC column and

RC slab

Plaster 15–20 mm thick each side

RC slab

Plaster 15–20 mm thick each side, existing minor cracks

walland timber RC shear wall wardrobe URM return walls both sides

RC slab

Plaster 15–20 mm thick each side, existing minor cracks

Features and preparations

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URM / RC

Plaster 15–20 mm thick each side

Plaster 15–20 mm thick each side, horizontal 50 mm deep cut at 1600 mm above floor height

WO1B

3900

2740

110

Timber (lateral only)

URM return walls both sides

URM / RC

WO1C

3900

2740

110

Timber (lateral only)

URM return walls both sides

URM / RC

Plaster 15–20 mm thick each side, horizontal and vertical 50 mm deep cut at 1600 mm above floor height and at the horizontal midway mark

WO2

2600

2740

110

Timber (overburden)

URM / RC

Plaster 15–20 mm thick each side, loadbearing wall (overburden load)

WR1

2180

4280

108

RC slab

URM return wall and short doorboth opening Free sides

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N A

RC slab

Side edges saw cut free

RC slab

WR2A tested in its existing condition prior to saw cutting edges and re-testing as WR2B

WR2A

2662

4342

108

RC slab

URM return wall and URM pier

WR2B

1915

4342

108

RC slab

Free both sides

RC slab

Side edges of WR2B saw cut free after testing WR2A

WR3

3385

2700

108

RC slab

Free both sides

RC slab

Side edges saw cut free (one side was saw cut above existing door opening)

WR4

1900

2450

108

RC slab

Free and tall door opening

RC slab

One side edge saw cut free and other side edge had nearly full-height door opening

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URM return wall and short door opening Free both sides Concrete-encased steel columns both sides

WR5

2580

2980

108

Timber (lateral only)

WR6

1305

2400

108

Timber (lateral only)

AG1

4400

3400

112.5

Shallow RC beam

AG2

4400

3400

112.5

Concrete-encased steel columns both sides

AO1

3380

2655

A

Shallow RC beam

109

RC beam

RC columns both sides

AK1

3350

2750

75

RC beam

RC column

AK2

1450

2750

75

RC beam

Free both sides

C C

RC slab RC slab

Side edges saw cut free

RC slab

Brick masonry veneer (as part of cavity infill wall) removed prior to testing

RC slab

Brick masonry veneer (as part of cavity infill wall) removed prior to testing, simulated in-plane cracking with 50 mm cut performed in a stair-stepped fashion along bed and head joints in X-shape across entire panel

Suspended timber, RCRC foundation beam Original cavity steel wire ties RC beam

Vertical through cut of the 75-mm brick and removed original cavity steel wire ties

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(b) Test wall WO1C after horizontal and vertical saw cuts 50 mm deep cuts horizontal at 1600 mm above wall base and vertical at wall centre

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(a) Test wall WO1B after horizontal saw cut 50 mm deep at approximately 1600 mm above wall base

Material properties

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(c) Test wall AG2 after 50 mm deep X-shaped cuts (d) Vertical through cuts for walls tested in performed in a stair-stepped fashion along bed and one-way flexure in the WR building head joints on wall loading side to simulate in-plane damage preceding OOP loading Figure 9. Test wall preparations

Brick, mortar, and masonry prism samples were extracted from the test walls and tested in accordance

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with ASTM standards as follows: brick compression strength, 𝑓𝑏′ (ASTM 2011a); mortar compression strength, 𝑓𝑗′ (NZSEE 2016); masonry prism compression strength, 𝑓𝑚′ (ASTM 2011b); ′ masonry prism bond rupture strength, 𝑓𝑓𝑏 (ASTM 2011c); and brick rupture strength / modulus of ′ rupture, 𝑓𝑚𝑟 (ASTM 2011a). The gross cross-section of bricks was assumed for determining all

material strengths. A summary of the material test results is included in Table 2, where all strength values are in units of MPa unless noted otherwise. Empirical equations were used to estimate the predicted mean values when it was not possible to test for certain material strengths as follows: 11

ACCEPTED MANUSCRIPT masonry prism compression strength (MPa), 𝑓𝑚′ = 0.75𝑓𝑏′

0.75

× 𝑓𝑗′

0.31

(Lumantarna et al. 2014);

′ masonry prism bond rupture strength (MPa), 𝑓𝑓𝑏 = 0.03𝑓𝑗′ (Almesfer et al. 2014); brick rupture ′ strength (MPa) / modulus of rupture (MPa), 𝑓𝑚𝑟 = 0.12𝑓𝑏′ (Almesfer et al. 2014); and masonry

prism density (kg/m3), 𝜌𝑚 = 1578 + 5𝑓𝑏′ + 8𝑓𝑗′ (Lumantarna 2012). For test walls WH1, WH2, and

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WH3, for which only masonry prism strengths were tested, the empirical equations were re-arranged to estimate values 𝑓𝑏′ and 𝑓𝑗′ . Material strength values assumed for test walls WR3 and WR4 were

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averaged from the corresponding values measured or estimated for tests walls WR5 and WR6, as all

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four test walls were in the same wing of the building and presumably constructed at approximately the same time.

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Knowledge of the masonry dimensions in test walls was required to compare the measured results to

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the predictive results of the AS (2011) method. The average measured brick height (mm), brick length (mm), and mortar joint thickness (mm) in each relevant building were as follows: 160, 300, and 15 for

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the WH building; 76, 230, and 18 for the WO building: 78, 223, and 13.5 for the WR building; and 76,

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225.5, and 13.5 for the AO building. Knowledge of the RC bounding element dimensions (as noted earlier) as well as expected concrete compression strength was required to compare the measured results to the predictive results of the Flanagan and Bennett (1999) method. The concrete compression

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strength for each relevant building was: 26 MPa for the WR building (Peng and McKenzie 2013);

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42 MPa for the AG building estimated as the specified strength for vintage concrete of 21 MPa (TNZ 2004) multiplied by 2.0 to account for age and overstrength (NZSEE 2006); and 34 MPa for the AO building estimated as the specified strength for vintage concrete of 17 MPa (TNZ 2004)) multiplied by 2.0 to account for age and overstrength (NZSEE 2006). Concrete compression strength of 28 MPa for the WH and AK buildings was estimated as the specified strength for contemporary concrete of 14 MPa (TNZ 2004) multiplied by 2.0 to account for age and overstrength (NZSEE 2006).

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Masonry prism bond ′ rupture strength, 𝑓𝑓𝑏

Brick rupture strength (modulus of rupture), ′ 𝑓𝑚𝑟

Masonry prism density, 𝜌𝑚 (kg/m3)

1.5

1650

0.50 4 18.7

0.26 3

Est.

Est.

Est.

Mean CV #

25.6 0.28 7

12.6 0.29 18

Mean CV # Mean CV # Mean

24.6 0.15 4 42.0 0.09 4 33.0

9.9 0.30 6 11.2 0.23 6 7.9

Avg.

Avg.

37.4 0.12 3 28.5 0.25

8.0 0.31 6 7.8 0.26

WR3 WR4

WR5

WR6

AG1 AG2

AO1

AK1 AK2

CV # Mean CV # Mean CV

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WR2

Est.

3.1

1807

Est.

Est.

0.30

3.0

1780

Est.

Est.

Est.

0.34

5.0

1878

Est.

Est.

Est.

Est.

19.5

0.24

4.0

1806

Avg.

Avg.

Avg.

Avg.

21.6

0.24

4.5

1829

Est.

Est.

Est.

Est.

17.5

0.23

3.4

1783

0.38

16.9 Est. 26.2

3 35.5 0.08

5 13.9 0.09

Est.

Est.

Est.

Est.

9.4 0.30

0.42

3.6 0.23

1720 0.03

#

5

5

2

4

3

Mean CV #

27.6 0.29 4

8.4 0.41 6

17.5

0.25

3.3

1783

Est.

Est.

Est.

Est.

Mean CV

8.0 0.27

1.2 0.22

3.8

0.04

1.0

1628 Est.

# Mean CV

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WR1

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WO1 WO2

Est.

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Est.

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CV #

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12.5

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Mean

D

WH1 WH2 WH3

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Masonry prism compression strength, 𝑓𝑚′

0.80

Brick compression strength, 𝑓𝑏′

13.8

Parameter

26.7

Test wall(s)

Mortar compression strength, 𝑓𝑗′

Table 2. Summary of measured and estimated masonry material characteristics (all strength values in MPa unless noted otherwise)

Est.

Est. Est. Est. # 7 13 CV = coefficient of variation defined as the sample standard deviation divided by the mean # = number of test samples Est. = estimated (predicted mean) value by empirical equation Avg. = average of corresponding WR5 and WR6 values

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3.2

Experimental test setup and instrumentation

Loading was applied to all test walls with an air compressor to gradually inflate between one and three (depending on the wall length) vinyl airbags that were positioned in a gap of 25 mm to 35 mm between the test wall panel and a plywood backing panel. The loaded area from each airbag was approximately 1150 mm by

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2050 mm. The plywood backing panel consisted of an assemblage of plywood sheets and timber frames [see Figure 10(a)–(c)]. The applied load from the airbags was transferred from the plywood backing panel to the

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braced reaction frame using between six and eight s-shaped load cells (each with a capacity of 10 kN), which

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provided the primary source of horizontal stability to the plywood-backed frame panel [see Figure 10(b)]. The

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plywood-backed frame panel rested on greased steel plates to allow the panel to slide with minimal frictional resistance [see Figure 10(b)] and ensure that the entire load was transferred through the load cells and not

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resisted by bearing friction. The braced reaction frame consisted of vertical and diagonal timber members screw fixed into the concrete floor slab [see Figure 10(c)]. The total lateral load, V, at any given time was

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calculated as the summation of the force recorded by all load cells.

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The instrumentation used to measure the OOP displacement of each test wall was generally placed on an isolated frame located on the opposite side of the test wall to the loading frame [see Figure 10(c)–(d)]. The

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instrumentation frame supported multiple strain gauges, string potentiometers, and LVDTs during any single test. Highly sensitive digital callipers were also placed redundant to other instrumentation at critical locations

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(i.e., on the wall mid-height) as back-up and to provide comparative measurements. A high-speed data acquisition (DAQ) system with multiple channels was used to record the test measurements at a frequency of at least 10 Hz.

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(b) Close-up of load cell placed between the plywood-backed panel frame and the reaction frame (top), and the plywood-backed frame atop greased low-friction steel plates

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(a) Reaction frame arranged with multiple air bags simultaneously inflated

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(c) Schematic of OOP test reaction frame (left side of wall cross-section) (d) Displacement instrumentation placed and displacement instrumentation (right side of wall cross-section) [h = on framing on the side of the wall test wall height, D = displacement gauge] (not to scale) opposite the reaction frame Figure 10. Test setup for OOP loading of wall panels

EARTHQUAKE DEMAND REQUIREMENTS AND PERFORMANCE CRITERIA An understanding of how the experimentally measured masonry wall OOP capacities compare to Design Basis Earthquake (DBE) demands in regions of varying seismicity is useful to structural engineering practitioners interpreting and applying the results of the experimental study reported 15

ACCEPTED MANUSCRIPT herein. Short-period (SDS at a period of 0.2 s) and long-period (SD1 at a period of 1.0 s) spectral accelerations for the DBE (assuming shallow subsoils) and relative levels of seismicity for the two regions in which the test buildings were located are as follows (NZS 2004; ASCE 2014): Wellington, New Zealand: SDS = 1.17 g; SD1 = 0.48 g; high seismicity

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Auckland, New Zealand: SDS = 0.38 g; SD1 = 0.15 g; moderate seismicity

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

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Because masonry infill walls are generally assumed to not contribute to the primary lateral load resisting system when loaded OOP, the earthquake demands assumed for the assessment of infill

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masonry walls acting OOP should be based on loading requirements for parts and components of

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buildings. Hence, the spectral demands prescribed for parts and components by the New Zealand earthquake loadings standard (NZS 2004) were considered in the study reported herein for the tested

TEST RESULTS AND DISCUSSION

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infill and partition walls in RC frame/slab buildings.

Measured and observed OOP wall performance

All test walls were laterally loaded semi-cyclically at a quasi-static loading rate. The maximum

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lateral-force value (expressed as an acceleration with respect to gravity, g) for each wall was

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determined by dividing the maximum total test lateral force, V, by the weight of the test wall [see Figure 11]. The peak test force is noted in Figure 11. In the cases of some extremely strong test walls, the peak test force was limited by the testing equipment. In many of the test buildings, the test walls were required to remain in place after testing. Hence testing was concluded after the peak strength of the test walls was reasonably assumed to have been reached. Idealised curves were added to the measured force-displacement curves shown in Figure 11 when peak wall strength was assumed to have been reached. The idealised curves shown in Figure 11 connect the origin, with the idealised 16

ACCEPTED MANUSCRIPT yield drift determined by assuming an equivalent elasto-plastic system with reduced stiffness (Park 1989) and the post-crack peak strength “design point”. Test wall AG2 was tested to complete collapse, and the instability drift was measured using photogrammetry. The relatively high lateral-force measurements at low drifts shown in Figure 11(p) represent situations where the test wall may have had small uncut portions of masonry at a boundary interface prior to that portion cracking. The values

PT

for OOP drift measured with respect to the initial base position are shown in Figure 11 as the ratio (%)

RI

of the OOP displacement at mid-height to the vertical distance between the wall base and the

0.00%

2.0

40 30

1.0

20

0.5

10 0

0.1

0.2 0.3 0.4 0.5 OOP displacement at midheight (mm)

0.6

0.7

PT E

0.0

D

0.0

OOP drift from initial base (%) 0.05% 0.10% 0.15% 0.14%

15

CE

3.0 2.5 2.0

1.5 1.0 0.5

25 20 10

0.5

5 0.0

0

0.0

0.5

1.0 1.5 2.0 2.5 OOP displacement at midheight (mm) Idealisation

3.0

Secant stiffness

(b) WH2

0.20%

0.18%, 46.8 kN

50 45 40 35 30 25

AC

V / (test wall weight) (g)

1.0

30

1.5

20

V (kN)

0.00%

45 35

2.0

Measured

(a) WH1

50 0.18%, 46.2 kN

0.14%

40

MA

1.5

V (kN)

50

2.5

0.20%

NU

60

2.5

V / (test wall weight) (g)

70

3.0 V / (test wall weight) (g)

3.0

OOP drift from initial base (%) 0.05% 0.10% 0.15%

V (kN)

OOP drift from initial base (%) 0.01% 0.02% 0.03%

0.00%

SC

mid-height displacement gauge (i.e., approximately half the wall height).

15 10 5

0.0

0

0.0

0.5

Measured

1.0 1.5 2.0 2.5 OOP displacement at midheight (mm) Idealisation

(c) WH3

3.0

Secant stiffness

(d) WO1A

17

PT

ACCEPTED MANUSCRIPT

(f) WO1C

(h) WR1

AC

CE

PT E

(g) WO2

D

MA

NU

SC

RI

(e) WO1B

(i) WR2A

(j) WR2B

18

PT

ACCEPTED MANUSCRIPT

(l) WR4

(n) WR6

AC

CE

PT E

(m) WR5

D

MA

NU

SC

RI

(k) WR3

(o) AG1

(p) AG2

19

PT

ACCEPTED MANUSCRIPT

0.29%, 19.9 kN 20

1.3

15

1.0 10

0.8

V (kN)

1.5

0.5

5

0.3

0.0

0

1.0 2.0 3.0 4.0 OOP displacement at midheight (mm) Idealisation

(r) AK1

Secant stiffness

D

Measured

5.0

PT E

0.0

2.0

OOP drift from initial base (%) 0.04% 0.06% 0.08% 0.05%

0.10% 12

0.08%, 10.2 kN 10

1.8

8

1.5 1.3

6

1.0

V (kN)

0.20%

0.02%

NU

1.8

V / (test wall weight) (g)

25

MA

V / (test wall weight) (g)

0.00% 2.3

0.30%

2.0

SC

OOP drift from initial base (%) 0.10% 0.20%

0.00%

(q) AO1

RI

(p) AG2 (showing instability displacement)

4

0.8

0.5

2

0.3

0.0

0 0.0 Measured

0.5 1.0 OOP displacement at midheight (mm) Idealisation

1.5

Secant stiffness

(s) AK2

4.1.1

WH Building

CE

Figure 11. Force-displacement responses for the test walls

AC

When testing WH1, the force measured by load cells was not recorded above a threshold of approximately 70 kN even when based on readings from an analogue pressure gauge; it was estimated that the testing continued until a maximum force of approximately 150 kN was reached. At the maximum loading pressure, WH1 showed no evidence of wall movement at the boundaries and no cracks were visually detected on the wall surface. The test results for WH2 are shown in Figure 11(a) in terms of equivalent acceleration at the wall base versus wall mid-height displacement. A crack pattern appeared to have existed before 20

ACCEPTED MANUSCRIPT commencement of the tests. Although monitoring the crack propagation was difficult due to a layer of paint on the wall, subsequent crack pattern observations confirmed that the wall damage remained limited to further opening of the existing cracks. At a force of approximately 25 kN, several vertical cracks were visible closer to the top (unrestrained) edge of the wall; thereafter, the crack pattern and crack widths remained constant.

PT

The lateral OOP response of WH3 is reported in Figure 11(c), which shows that at a loading of

RI

approximately 47 kN, a wall stiffness reduction of approximately 66% (from 30 kN/mm to 20 kN/mm)

SC

was observed. WH2 and WH3 exhibited small amounts of plastic deformation (approximatively 0.5 mm), while WH1 behaved elastically throughout testing. Each of the walls tested in the WH building

WO Building

MA

4.1.2

NU

satisfied the current seismic code demand pressure (NZS 2004).

The results obtained for WO1 before the introduction of additional cuts (tests WO1B and WO1C) to

D

simulate progressive damage are reported in Figure 11(d), which plots the displacement measured at

PT E

the centre of the panel. The backbone curve shows relatively linear behaviour (secant stiffness approximately equal to 21 kN/mm) with small residual displacements up to pressure values higher than those corresponding to a ground acceleration of 0.8 g. The load was applied through a series of 0.5 kPa

CE

incremental steps. No cracking was observed on the side opposite to the loaded face (tension side) of

AC

the walls, which was attributed to a painted surface. The vertical displacement profile highlights that the constraint at the top edge of the wall proved ineffective in restraining the horizontal displacement of the upper part of the panel. The horizontal displacement profile shows a symmetrical deflection about the centre line of the wall. As previously described, a horizontal 50-mm-deep cut was introduced at two-thirds of the wall height to simulate a damaged state due to past earthquake events [see Figure 9(a)]. The backbone curve for WO1B (Figure 11(e)) shows a relatively linear relationship up to lateral loading of 25 kN, with slightly reduced secant stiffness (20 kN/mm). For loading pressure higher than 21

ACCEPTED MANUSCRIPT 25 kN, the wall stiffness appeared to be significantly lower, with a backbone curve slope of 9.9 kN/mm. The OOP capacity of the wall was then further diminished by the introduction of a 50-mm-deep vertical cut [Figure 9(b)] at the centre of the wall section. Figure 11(f) shows a response characterised by lower stiffness (16.4 kN/mm) and a strength plateau at approximately 47.8 kN. The presence of a door opening in WO2 resulted in reduced stiffness compared with wall section WO1,

PT

where effective lateral boundary constraint on both sides was present. The backbone curve of the WO2

RI

test [Figure 11(g)] shows that the wall had an initial linear branch with a stiffness of 25.3 kN/mm. For

SC

lateral loads greater than 19 kN, the response envelope indicated that wall secant stiffness was reduced by approximately 50% to 12.6 kN/mm. WR Building

NU

4.1.3

MA

The test results for WR1-WR6 are reported in Figure 11(h)-(n). WR1 and WR2B were of similar size and had a similar response, reaching a strength plateau at similar force levels and displacements at the

D

centre of the panel (> 10 mm). It is evident from Figure 11 (where WR2A is compared with WR2B)

PT E

that the removal of lateral vertical constraints by cutting WR2 and consequently forcing it to respond with single flexure resulted in a significant decrease in both stiffness and capacity. The maximum loading pressure applied to WR2A is related to the maximum capacity of the reacting frame. Such

CE

pressure was reached with a displacement of 1.8 mm at the centre point of the wall panel and hence it

AC

was possible to perform only a limited number of cycles (minimum displacement-controlled steps of approximately 0.5 mm). Tests WR3 and WR4 [Figure 11(k, l) respectively] were performed on walls with one-way vertical bending capacity, similar to WR1 and WR2B, but with shorter wall-panel height. This reduced height resulted in higher initial stiffness and a less pronounced non-linear response. The effect of unbounded restraint conditions (i.e., no stiff frame or slab member in a considered direction of flexure) was investigated for two-way flexure in test WR5 [Figure 11(m)] and one-way flexure in test WR6 [Figure 11(n)]. It was observed that unbounded conditions resulted in 22

ACCEPTED MANUSCRIPT more non-linear responses, larger residual displacements, and significant reductions in stiffness and strength relative to bounded conditions. 4.1.4

AG Building

Test AG1 [Figure 11o] was carried out in two-way bending flexure after the removal of the external

PT

masonry leaf. The panel reached capacity (61.3 kN) at a relatively large mid-height displacement (approximately 3.0% drift). No substantial strength loss was observed up to the critical mid-span

RI

displacement, which was reached at approximately 100 mm. Multiple step cracks [Figure 12(a)]

SC

originated from the wall centroid and propagated toward the panel edges without exhibiting the typical central horizontal crack that is commonly observed for rectangular, two-way spanning walls (Griffith

NU

& Vaculik [2007]). AG2 was tested subsequent to the introduction of an X-shaped cut aimed at

MA

reproducing the cracking pattern of a wall that had experienced substantial in-plane loading [see Figure 9(c)]. AG2 exhibited critical mid-height displacement at the point preceding the collapse

D

[Figure 12(c)] at approximately 120 mm. The in-plane simulated state of damage had a detrimental

PT E

effect on wall capacity (38.4 kN), which decreased approximately 40% compared to AG1 [see Figure 11(p)]. The presence of the X-shaped simulated in-plane cracks also limited the formation of other

AC

CE

cracks by acting as a pre-defined diagonal pattern [Figure 12(b, c)].

(a) Crack pattern of AG1

(b) Crack pattern of AG2

(c) Ultimate failure of AG2

Figure 12: Crack patterns observed for tests in the AG building

23

ACCEPTED MANUSCRIPT 4.1.5

AO Building

Results for test AO1 are shown in Figure 11(q). A maximum possible load of 63.9 kN was applied (due to test setup limitations) with 0.1% drift attained. AO1 exhibited considerable initial stiffness at the mid-height of the wall, and no damage to the wall was observed. AK Building

PT

4.1.6

Results for tests AK1 and AK2 are shown in Figure 11(r, s), respectively. For AK1 a maximum load

RI

of 19.9 kN was applied with 0.29% drift attained, and for AK2 a maximum load of 10.2 kN was

SC

applied with 0.08% drift attained. Due to its single-leaf thickness, the removal of all steel tie connections, and the different side boundary when compared to AK1, test AK2 resulted in higher OOP

NU

displacement when subjected to similar load magnitudes than did test AK1 for the two-leaf cavity wall.

4.2

MA

No cracking was observed in any of the tested walls.

Conversion of measured forces to equivalent full-height, uniformly loaded OOP capacities

D

For the purpose of interpreting the measured behaviour from OOP proof tests, the measured peak

PT E

total force on any given test was converted into an equivalent full-height, uniformly loaded OOP capacity for each wall to account for the following issues: The loaded wall area was often smaller than the total wall area;



The airbags were not always perfectly vertically centred on the wall; and



For each test wall prepared with vertical saw cuts to free the side (vertical) edges, the tested

AC

CE



wall height (as listed in Table 1) may have been shorter than the full in-situ wall height due to spatial restrictions on saw access, as was the case with WR3, WR4, and WR6. In such an instance, the test wall may have behaved diffrently during testing due to a lower height-to-thickness slenderness ratio than would have been present under actual earthquake loading (notwithstanding the inherent conservative nature of testing walls in one-way flexure). For walls tested in two-way flexure, due to boundary restraints at the top or bottom as well as on at 24

ACCEPTED MANUSCRIPT least one side (vertical) edge (as noted in Table 1), a simplified approach was adopted for converting the measured test forces into fully distributed force capacities. Hence, for the walls listed in Table 3 that were tested in two-way flexure with full in-situ dimensions, the determined force-based capacity was simply derived by dividing the maximum total test lateral force, V [see Figure 11], by the in-situ wall dimensions, resulting in a conservative determination of uniformly loaded OOP capacity provided

PT

that the airbags were approximately centred on the wall (as was the case in the tests reported herein).

RI

For walls tested in one-way flexure, due to saw cutting on both sides (vertical edges), an OOP force

SC

conversion process proposed by Walsh et al. (2015) was applied. The results of this conversion from the test scenario to the assessment scenario are summarised in Table 3 for test walls WR1, WR2B,

NU

WR3, WR4, WR6, and AK2. The assumed point of equivalence in the analytical conversion from the test scenario to the assessment scenario is the maximum flexural capacity of each wall at the

MA

cross-section corresponding to the primary horizontal crack occurring at a height above the wall base, represented by the crack height ratio (i.e., the vertical distance of the primary crack above the test wall

D

base divided by the test wall height), which was maintained as equal for both conditions. For this

PT E

conversion, all walls were considered to be simply supported. This analytical conversion approach produces results similar to a conversion procedure incorporating external virtual work with a constant

CE

unit displacement occurring at the crack location, such as that utilised by Angel et al. (1994). However, the analytical conversion procedure proposed by Walsh et al. (2015) and applied in the testing

AC

programme reported herein more readily accommodates varying wall heights (i.e., test wall height versus in-situ wall height) and eccentric loading locations. While not considered in the study reported herein, multiple analytical models exist for predicting the strength and drift/displacement performance of walls under OOP loads (e.g., Abrams et al. 1996; Flanagan and Bennett 1999; Doherty et al. 2002; AS 2011; Derkahshan et al 2014b; Vaculik and Griffith 2017). These predictive models are considered in greater detail in a companion article (Walsh et al. 2018). Additional predictive models and reliability comparisons pertaining to the OOP 25

ACCEPTED MANUSCRIPT performance of URM walls have been considered by others (e.g., Komaraneni et al. 2011; Mosalam and Günay 2015; Furtado et al. 2016; Libratore et al. 2016; Shing et al. 2016; Asteris et al. 2017; Pasca

AC

CE

PT E

D

MA

NU

SC

RI

PT

et al. 2017).

26

ACCEPTED MANUSCRIPT

Test ID

Vertical saw cuts to free both side (vertical) edges?

Table 3. Summary of conversion of measured loads from the test condition to the equivalent seismic condition

WH1

No

72.7

-

-

-

-

WH2

No

46.2

-

-

-

-

WH3

No

46.8

-

-

-

WO1A

No

23.6

-

-

-

WO1B

No

39.7

-

-

WO1C

No

47.8

-

-

WO2

No

22.9

-

-

WR1

Yes

16.1

7.9

WR2A

No

41.6

-

WR2B

Yes

13.5

WR3

Yes

40.3

WR4

Yes

42.1

WR5

No

65.1

WR6

Yes

7.6

AG1

No

AG2

No

AO1

No

AK1 AK2

full-height wall (g)

Equivalent force-based capacity of

Weight of wall full-height (kN)

PT

distributed load (kN/m)

Equivalent full-height uniformly

Full in-situ height of wall (mm)

SC

RI

(kNm/m)

Max flexural moment at crack height

-

22.7

3.20

2730

-

16.2

2.86

-

2730

-

14.6

3.20

-

2740

-

20.8

1.13

-

-

2740

-

20.8

1.91

-

-

2740

-

20.8

2.30

-

-

2740

-

13.9

1.65

2300

0.54

6.0

4280

5.7

17.6

1.39***

-

-

-

4342

-

23.0

1.81

6.6

2090

0.48

5.9

4342

4.8

16.5

1.25***

19.7

1350*

0.50*

5.0

3100**

14.1

20.1

2.17***

20.5

1460

0.60

7.6

3100**

12.5

11.3

3.43***

-

-

-

-

2980

-

14.9

4.37

3.7

1470

0.61

1.9

2980**

2.4

7.3

0.96***

61.3

-

-

-

-

3400

-

28.4

2.16

38.4

-

-

-

-

3400

-

28.4

1.35

63.9

-

-

-

-

2655

-

17.1

3.74

No

19.9

-

-

-

-

2750

-

11.0

1.81

Yes

10.2

5.0

1375

0.50*

3.0

2750

4.7

4.8

2.68***

D

PT E

CE

AC

MA

3600

NU

Crack height ratio

(mm)

load (kN/m)

Crack height above test wall base

full-height, uniform forces

Partial-height uniformly distributed

with partially distributed test forces

(kN)

Equivalent full-height walls with

Maximum total test lateral force, V

Test walls at partial, saw-cut height, and

*Assumed mid-height cracking for purposes of the force capacity conversion procedure. No primary cracking was actually observed during testing. **The full in-situ wall height is taller than the test wall height listed in Table 1. *** Equivalent force-based capacity of full-height wall (g) different than the measured test value shown in Figure 11.

27

ACCEPTED MANUSCRIPT Recall that the values for OOP drift from the initial base shown in Figure 11 were measured as the ratio (%) of the OOP displacement at mid-height to the vertical distance between the wall base and the mid-height displacement gauge (i.e., approximately half the wall height). The more traditional definition of OOP drift for the purpose of assessing URM walls is the ratio of OOP displacement measured at mid-height to the entire wall height (Flanagan and Bennett 1999; ASCE 2014), resulting

PT

in drift values that are approximately half as large as those reported in the study herein. The choice to

RI

consider drift over only half the wall height in the study reported herein was motivated by the desire to

SC

better align the experimental results with existing predictive models in which the OOP displacement capacity is measured at the height of the primary crack effectuating OOP collapse (generally near the

NU

mid-height of the wall), and where the height of the primary crack affects the predicted OOP strength and displacement capacity of the wall (e.g., Derakhshan et al. 2014b). All else equal, lower primary

MA

crack heights [see Table 3] are associated with more desirable collapse mechanisms (i.e., walls with low primary crack heights do not collapse as far from their initial vertical planes) and correspondingly

D

higher drift capacities, assuming that drift is measured over the vertical distance between the wall base

PT E

and the primary crack location. In a wall restrained at the top and bottom by relatively rigid supports, the compressive strut forces on the wall from arching action are likely to be far larger than the gravity

CE

loads from self-weight. As a result, primary horizontal cracking for walls tested quasi-statically in one-way vertical flexure with arching action is expected to occur at approximately half of the total wall

AC

height (Abrams et al. 1996) rather than approximately two-thirds of the total wall height as is typical for walls tested without arching action (Derakhshan et al. 2014b; Walsh et al. 2015). 4.3

Comparison of measured wall capacities to demand requirements

The estimated capacity/demand (C/D) ratios for four scenarios are summarised in Table 4 for the walls tested in the experimental programme reported herein. Demand accelerations listed in Table 4 were based on the DBE for the ultimate limit state (ULS), which is theoretically equivalent to the life safety (LS) performance level considered in ASCE (2014), for each of the two cities in which the test 28

ACCEPTED MANUSCRIPT buildings are located (i.e., Wellington and Auckland). The natural period of each uncracked wall, Tp, was conservatively assumed to be less than 0.50 s such that the part spectral shape coefficient, Ci(Tp), is 2.0 for walls assessed using the NZS (2004) demand requirements. Note that Walsh et al. (2015) determined that the average calculated natural periods for cavity walls tested in one-way flexure with RC top restraints and timber (lateral only) top restraints are 0.40 s and 0.90 s, respectively. Assuming a

PT

shallow subsoil site class, non-ductile OOP behaviour of the wall (which is appropriate for a peak

RI

force-based assessment), a part risk factor, Rp = 1.0, a building importance level of 2 (representing a

SC

normal building and hence warranting the consideration of a DBE with an average return period of 1 in 500 years), and in-situ full-height wall geometries and densities, the C/D ratio for each of the test walls

AC

CE

PT E

D

MA

NU

was estimated for each of four scenarios summarised in Table 4.

29

ACCEPTED MANUSCRIPT Table 4. Summary of capacity/demand (C/D) ratios considering equivalent full-height loaded capacities and assuming sites with shallow subsoils Capacity/demand (C/D) ratios based on force-based capacity and “parts and components” demands Wellington, ground floor of 4.5 m tall building Seismic Demand (g)

C/D

Wellington, third floor of 12 m tall building Seismic Demand (g)

C/D

Auckland, ground floor of 4.5 m tall building Seismic Demand (g)

C/D

Auckland, third floor of 12 m tall building Seismic Demand (g)

PT

Test ID

C/D

2.98

108%

0.45

713%

0.97

331%

WH2

1.31

219%

2.90

98%

0.42

673%

0.94

303%

WH3

1.31

245%

2.90

110%

0.42

755%

340%

WO1A

1.31

87%

2.90

39%

0.42

267%

0.94

120%

WO1B

1.31

146%

2.90

66%

WO1C

1.31

176%

2.90

79%

WO2

1.31

126%

2.90

57%

WR1

1.44

96%

3.04

WR2A

1.45

125%

3.04

WR2B

1.45

86%

3.04

WR3

1.34

162%

WR4

1.34

WR5

0.42

449%

0.94

202%

0.42

541%

0.94

243%

0.42

388%

0.94

175%

46%

0.47

296%

0.99

141%

59%

0.47

385%

0.99

183%

41%

0.47

265%

0.99

126%

2.93

74%

0.44

499%

0.95

228%

256%

2.93

117%

0.44

788%

0.95

360%

1.33

329%

2.92

150%

0.43

1013%

0.95

460%

WR6

1.33

72%

2.92

33%

0.43

222%

0.95

101%

AG1

1.37

158%

2.96

73%

0.44

486%

0.96

224%

AG2

1.37

99%

2.96

46%

0.44

305%

0.96

141%

AO1

1.30

288%

2.90

129%

0.42

885%

0.94

397%

1.31

138%

2.90

62%

0.43

425%

0.94

192%

1.31

205%

2.90

92%

0.43

631%

0.94

284%

AK1

NU

MA

AC

AK1

SC

0.94

CE

RI

232%

D

1.38

PT E

WH1

In accordance with the values shown in Table 4, most of the tested URM walls, especially those with RC boundary restraints or boundary restraints effectuating two-way flexure OOP, were generally expected to perform well compared to DBE demands in areas of moderate seismicity (e.g., Auckland). However, when subjected to higher spectral demands by the DBE in areas of high seismicity (e.g., Wellington), especially when located at higher storeys and subjected to higher spectral demands, many of the tested walls were deemed to be at risk due to their maximum OOP strength being exceeded. 30

ACCEPTED MANUSCRIPT Note that the C/D ratios listed in Table 4 should be strictly considered in relation to the geometry and material properties of the walls tested in the experimental programme reported herein. However, an inherent conservativeness exists within both force-based assessments and one-way vertical flexural analyses. SUMMARY AND CONCLUSIONS

PT

5

RI

A total of 19 URM infill walls located in six buildings in New Zealand were tested for OOP

SC

behaviour. The testing procedures, measured material strength, recorded lateral wall behaviour, and observed wall damage are presented and discussed. The following significant results can be drawn



NU

from this research program:

The testing showed that the walls are capable of resisting seismic demands in regions with

MA

moderate to high seismicity (e.g., Wellington, New Zealand) despite some simplified predictive methods suggesting lower strengths for some walls. Considered predictive models

Restraint at the walls’ vertical edges (horizontal boundaries), resulting in two-way OOP

PT E



D

are applied in greater detail in a companion article (Walsh et al. 2018);

flexure as compared to one-way vertical OOP flexure, can substantially improve the OOP

Topside fixed restraint and presumed ‘arching’ action from the building frame can greatly

AC



CE

load-carrying capacity of tested infill walls;

increase the OOP capacity of URM walls; 

In-plane damage was found to reduce the OOP capacity of URM infill walls by up to 40%;



Material strength related to brick compression, mortar compression, masonry bed joint shear, and cavity tie pull-out, as well as other properties, were determined for a range of buildings in this typology, providing evidence for engineering consultants on the merits of conducting site investigations to make more accurate assumptions when performing future building analyses; 31

ACCEPTED MANUSCRIPT 

On-site proof testing was shown to be a simple and cost-effective way to establish actual lateral capacities of walls in cases where boundary conditions cannot be clearly established and the models predict low lateral capacities.

Further research in this area should involve advanced data processing to define more accurately the

PT

effects of different boundary conditions on URM infill wall performance. Further testing will preferably involve an examination of retrofit techniques. Some walls, for example, may be retrofitted

RI

with vertical near-surface-mounted carbon fibre strips, which is a cost-effective and

SC

minimally-invasive seismic retrofit technique for some scenarios where cavity ties are not appropriate,

ACKNOWLEDGEMENTS

MA

6

NU

particularly for walls that are propped at the top.

A portion of the experimental testing was funded by the Building Research Association of New

D

Zealand (BRANZ) through grant LR0441 and the Natural Hazards Research Management Platform

PT E

(NHRP) through grant C05X0907. The authors are also grateful for the in-kind donations provided by the owners of the tested buildings, including KiwiRail and Mansons TCLM Ltd. Technical advisory

CE

for both testing and analysis was provided by Holmes Consulting Group. Students and staff who participated in the various field and laboratory testing efforts include Anthony Adams, Mark Byrami,

AC

Marta Giaretton, Mark Liew, Jeff Melster, Alexandre Perrin, Laura Putri, Jerome Quenneville, Ross Reichardt, and Gye Simkin.

32

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