Prefabricated connection for steel beam and concrete-filled steel tube column

Journal of Constructional Steel Research 162 (2019) 105751

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Journal of Constructional Steel Research

Prefabricated connection for steel beam and concrete-filled steel tube column Chenting Ding a, Xuebei Pan b, Yu Bai a,⁎, Gang Shi c a b c

Department of Civil Engineering, Monash University, Clayton, VIC 3800, Australia School of Civil Engineering, Hefei University of Technology, Anhui Province 230009, China Key Laboratory of Civil Engineering Safety and Durability of China Education Ministry, Department of Civil Engineering, Tsinghua University, Beijing 100084, China

a r t i c l e

i n f o

Article history: Received 30 March 2019 Received in revised form 11 August 2019 Accepted 23 August 2019 Available online xxxx Keywords: Concrete-filled steel tube Steel structures Beam-column connection Prefabrication Bolted connection Moment-rotation behaviour

a b s t r a c t A new type of connection system with steel beam and concrete-filled steel tube (CFST) column is proposed in this paper. This type of connection system uses internal stiffeners and high-strength threaded steel rods to connect the steel beams in two directions with the assistance of side plates pre-welded on the column, intending to serve for high-rise buildings where moderate or high ductility is required, and is geometrically possible to be covered within the wall or floor space therefore offering architectural pleasing. The force transfer mechanisms of the connection are important for the corresponding load-carrying capacities and may be different in the two beam directions. In order to investigate the mechanical properties of the connection system, four different specimens, including interior and exterior connections with corresponding columns and beams in two directions, were made and examined under monotonic loadings, further with comparison to FE modelling results. The results of experimental and numerical investigations indicate the proposed connection system offers satisfactory stiffness and load-carrying capacities. Force transfer mechanisms are discussed with assistance from stress and strain responses identified from validated FE modelling in details. Design considerations are further highlighted for implementation in practice. © 2019 Elsevier Ltd. All rights reserved.

1. Introduction Due to their ductility and high capacity, concrete-filled steel tubes (CFSTs) are commonly used as columns in building structures. The infilled concrete is confined by the steel tube, resulting in a triaxial state of compression whereby both load capacity and ductility are improved. Axial compression behaviours [1–3] and bending behaviours [4–6] of CFST columns were widely studied and all the evidences indicate that both compressive and bending capacities of CFST columns are high. One of the challenges for implementation of CFST members in building structures is the development of proper connection approaches. Investigations of connections between steel beams and CFST columns have therefore received important attention. Generally, such connection specimens for experimental studies are vertically loaded at beam ends and are constrained at the top and bottom ends of the column, and the deformation capacities of connections can be quantified through equivalent storey drift angle [7]. Such steel beam to CFST column connections can be categorized into two groups [8]: connections that attach only to the steel tube of the CFST column, and connections ⁎ Corresponding author. E-mail addresses: [email protected] (C. Ding), [email protected] (Y. Bai).

https://doi.org/10.1016/j.jcsr.2019.105751 0143-974X/© 2019 Elsevier Ltd. All rights reserved.

that also have elements embedded in the concrete core of the CFST column. For the former, usually no proper measure is employed to stiffen the panel zone of the joint. For example, blind bolts were proposed for extended end-plate connections and flush end-plate connections to link end plates to the steel tube, as investigated in [9,10]. The experimental results indicated that an increased end plate thickness can evidently enhance the stiffness and reduce the rotation capacity of these two kinds of connections. Alternatively, the flanges of steel beams can be directly welded to the surface of the CFST column. As found in experimental and numerical investigations conducted in [8,11], this type of connections was associated with low flexural strength due to cracked welds. Overall, the two design methodologies of connections ([9,11]) did not employ any measure to stiffen the panel zone. Therefore, the column steel surface could be bent outward by the beam flanges through tension, resulting a relatively low moment-rotation stiffness. Connections with these design methodologies can be classified as semi-rigid connections according to the experimental results for the initial elastic stiffness. From this point of view, moment-rotation stiffness and capacity might be improved if the joint panel zone is stiffened. A practical way to stiffen the panel zone is through the use of external diaphragms as stiffening members; this method has been widely considered in engineering practice [12]. An external diaphragm is an annular plate

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surrounding the column and welded to the steel tube surface. A connection may consist of 2 external diaphragms, normally corresponding to the upper and lower flanges of the steel beam (Fig. 1a). Previous studies [8,13–15] have shown that properly designed connections with external stiffeners could offer improved moment-rotation stiffness and capacity. For example, in [8], a connection with external diaphragms offered 18% higher initial stiffness and 20% greater capacity than a connection of the same size but without external diaphragms, that is, where the beam was directly welded to the steel tube. A noticeable advantage of this kind of connection is that concrete can be poured into the column tube conveniently because no other structural components are inside the steel tube. However, because the diaphragm stiffeners occupy external space and cannot be fully concealed within the wall or floor, their visibility within the indoor space may not be aesthetically pleasing. The joint panel zone can be stiffened by structural elements embedded in the concrete core. An internal diaphragm (Fig. 1b), as a horizontal steel plate within the column, welded to the interior steel surface and with a hole normally at the centre for concrete casting, has been used to stiffen joint panel zones [12]. Again, such a connection may include 2 internal diaphragms, corresponding to the upper and lower flanges of the steel beam. The shear capacity of steel beam to CFST column connections stiffened by internal stiffeners was investigated in [16]. It was found that the shear strength of the panel zone stiffened by internal stiffeners was about 130% of the design requriement and the column steel surface was not obviously bent outward. A theoretical study on the multi-linear load-deformation relations of the internal diaphragm connection between plate and tubular column with concrete filling using yield line method was provided specifically by Fukumoto [17,18]. Further experimental studies [19] showed that connections with internal diaphragms could provide high moment-rotation stiffness, together with satisfactory moment capacity that allowed beams to fail prior to the joint. Shear behaviour of the joint panel zone was studied in [20] and such zones stiffened by internal diaphragms were also studied in [21], where the shear force and shear deformation relationship was formulated for the panel zone. Compared to the external diaphragms, internal diaphragms are in the column steel tube and surrounded by concrete. Therefore, anti-rust processing may not be necessary for internal diaphragms. However, it should be noted that in such cases the internal diaphragm often has a large hole in the centre for concrete pouring, therefore requiring a large internal diaphragm plate. Consequently, large cross-sections of the steel tubes in CFST columns are also needed. As well, welding of diaphragms inside a steel tube is laborious. Another way to enhance joint stiffness is to situate the steel beam through the steel tube so that the beam section can stiffen the joint

panel zone. This configuration requires making I-shaped holes (for the I beam section) in the steel tube of the CFST column. A series of experiments performed in [22] investigated the ductility of such connections. It was found that when the design principle of strong column and weak beam was considered, the connections were very ductile and the equivalent drift angle could reach 9.6%. A design guideline for this kind of connection was developed in [23] based on theoretical, numerical and experimental analyses. It was recommended that the column-to-beam strength ratio should be greater than 1.5 for connections in which full penetration welds were used to join the beam to the column, or should be greater than 2 for connections in which fillet welds were used. Again, the beam passing through the column may affect concrete pouring through the steel tube. Therefore, threaded steel rods were proposed to pass through the steel tube and concrete core in the panel zone of the CFST column. The steel rods were embedded in the concrete and extended to the outer surfaces of the steel tube, and were bolted to 2 angle plates (Fig. 1c) on the steel beams on two sides of the column. Such connections were developed for the steel beam to CFST column connection where threaded steel rods were placed to pass through the column in [24]. Cyclic bending loads were applied to the connections and experimental results indicated that such connections offered satisfactory deformation capacities with equivalent storey drift angles greater than 6%. Another steel beam to CFST column connection was examined in [25], where the column steel tube was made with I-shaped holes for the steel beam and with circular holes for threaded steel rods to pass through the panel zone (Fig. 1c). Experimental studies were conducted for such connections subjected to bending moments; the results showed that they had good ductility and could reach an equivalent storey drift angle of more than 5%. It should also be noted that such penetrating steel rods were also used for connection systems between CFST columns and reinforced concrete beams, in which steel bars passed through the connection horizontally without a steel tube in this region, while lateral hoop reinforcements were required to confine the concrete core in the connection zone to compensate for the interruption of the steel tube [26,27]. Several configurations have been developed for the connections between steel beams and CFST columns. It appears that connections with stiffened panel zones may offer better moment rotational stiffness and capacity than those without. However, current practice for such connections with internal or external diaphragms requires a large amount of site work including full penetration welding or the inconvenience of concrete casting. Also, from architectural point of view, it is desirable to have columns concealed within walls. In this study, a new connection system (Fig. 2) is proposed that considers the requirements of high moment-rotation stiffness, rapid on-site installation and architectural

Fig. 1. Illustrations of various types of beam to CFST connections (a) with external diaphragm configuration; (b) with internal diaphragm configuration; and (c) with through beam or steel rods.

C. Ding et al. / Journal of Constructional Steel Research 162 (2019) 105751

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Fig. 2. Proposed beam–CFST column connection.

appeal. The design concept and configuration in detail are introduced in the following. Furthermore, 4 specimens of the new connection system were prepared and were tested under static bending. The experimental results are presented and discussed in this paper. Detailed FE analyses were conducted and the modelling results are compared and verified against experiments. Based on the experiments and FE analyses of the proposed connection, the stiffness, ductility of connection, and forces distributed in stiffening members are studied thoroughly. 2. Design concept and configuration In the proposed connection system, no internal diaphragm or external diaphragm is employed. Instead, the panel zone is stiffened by 2 internal stiffeners in the Y direction (shown in Fig. 2) and 4 highstrength circular threaded steel rods in the X direction (where the Y direction is the weak axis direction for the column section while X is the strong direction). An internal stiffener offers greater bending and tensile stiffness than 2 steel rods, due to the inertial moment of its cross-section. Therefore, the internal stiffener is used in the Y direction. Conventional horizontal diaphragms, including internal and external diaphragms, are replaced by internal vertical stiffeners, thereby offering more space in the column steel tube for concrete casting. In the other direction, threaded rods are made of highstrength steel, making them strong enough for axial tension forces. Both stiffening member types do not require large internal space or larger steel tubes, which enables the column section to be relatively small or identical to the wall thickness, permitting installation within the wall, and thereby offering architectural appeal. In this connection system, beams are not welded to the column but are bolted to the column with the assistance of side plates or flange cleats (see Fig. 2). Therefore, on-site welding work is reduced. Fig. 3 shows the fabrication and installation process. In the factory, 4 rectangular holes are drilled in the column for internal stiffeners to pass through the column, and 8 circular holes or slotted holes are

drilled in the column for threaded steel rods to pass through the column. Then, 2 internal stiffeners are inserted into the column through the rectangular holes and are welded to the column by fillet welds (see Fig. 3a). All the lower flange cleats, lower side plates, and fin plates are welded to the column by fillet welds. Next, the column with all the welded components is transported to the construction site. On site, the steel beam can be installed by joining the lower flange cleats, lower side plates and fin plates using high-strength bolts as shown in Fig. 2. Subsequently, the upper side plates and flange cleats are welded to the column by fillet welds as indicated in Fig. 3b, and are then bolted to the beam end using high-strength bolts. In this way, the steel beam can be placed in position easily using the prefabricated lower side and fin plates and the connection approach can also tolerate geometrical inaccuracy or imperfection through the on-site upper side and fin plates. Compared with the installation processes of conventional internal or external diaphragms, the installation process of this connection system requires much less on-site welding work and therefore may be faster because full penetration welds are not involved. In the next step, steel rods are passed through the column and tightened on both sides. Finally, selfcompacting concrete is cast into the column. 3. Experimental program 3.1. Specimens and material The experimental investigation considers two variables. One is the bending direction, i.e. about the strong axis (X axis) or weak axis (Y axis) of the column section, as the moment-rotation behaviours of the connection system along different directions can be different because the configuration of threaded steel rods is in X direction while that of internal stiffeners is in Y direction. The second variable is the location of the connection, i.e. with an interior column inside a building, or an exterior column at the periphery of a building. Four specimens were

Fig. 3. Installation process (a) components are welded to column before transport to construction site; (b) on-site beam installation and fillet welds; and (c) steel rods installation.

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therefore designed and named as SX1, SY1, SX2 and SY2. The beginning letter “S” denotes the specimen for experimental investigation rather than FE modelling. The second letter “X” or “Y” means the direction of bending (see Fig. 2a). X or Y direction is the strong or weak axis of the column in bending. The number “1” means the connection for interior column and “2” for exterior column. In a specimen for the connection with an interior column (SX1 or SY1), two beams with I-shaped sections (Fig. 4) in either X or Y direction were installed to the column side face. In a specimen for exterior connection (SX2 and SY2), only one beam with I-shaped sections (Fig. 4) in either X or Y direction were installed to the column side face. To allow the column to be concealed within a wall, the column has a rectangular section with the side length of 440 mm in the X direction and 220 mm in the Y direction (see Fig. 4). Moreover, the beam in specimen SY2 is eccentrically attached to the column and aligned to the column edge so that the column is invisible outside the wall. The eccentricity (distance between the centreline of the beam and the centre of the column) is 110 mm. It should be noted that when the beam is loaded with bending moment, the column is subject to torque induced by the eccentricity. The height of the column in each specimen is 2.82 m and the span length from the beam end to the centre of the column in each specimen is calculated to be 2.53 m. It should be noted that the height and length are determined to introduce the ends of the beams and columns of the specimens as inflection points of corresponding members in a potential building prototype. On that basis, the equivalent storey drift angle θd can be calculated by Eq. (1) [7], in which Δ1 and Δ2 are the vertical displacements at the ends of 2 beams for specimens SX1 and SY1. In specimens SX2 and SY2, Δ1 is the vertical displacement of the beam end and Δ2 equals zero. Δ3 is the horizontal displacement at the top of the column and Δ4 is the horizontal displacement at the bottom of the column; Lbeam is the span length of the beams connected to the column and Hcolumn is the height of the column. θd ¼

Δ1 −Δ2 Δ3 −Δ4 − Lbeam Hcolumn

ð1Þ

It should be noted that the thickness of each column is 220 mm and is close to the thickness of common walls. The dimensions of each side plate are 250 × 170 × 10 mm. The dimensions of each flange cleat are

shown in Fig. 4a and the thickness of the stiffener is 30 mm. The dimensions of each fin plate for the connection between the column and the beam web are 260 × 105 × 10 mm. The dimensions of each internal stiffener for the connection in the Y direction are shown in Fig. 4b and its thickness is 30 mm. All the steel beams, column steel tubes, flange cleats, side plates and internal stiffeners were made of Q345 steel (with the nominal tensile yielding capacity of 345 MPa). The concrete inside the column was C50 (with the nominal compressive capacity of 50 MPa). Three 150 mm concrete cubes were cast and cured in conditions similar to those of the specimens, with the average measured cube strength of 52 MPa. The bolts to connect the 2 beam flanges with the flange cleat or side plate were Grade 10.9 M22, with the exterior diameters of the bolts being 22 mm. The bolts used to join beam webs to the fin plates were Grade 10.9 M20. The threaded steel rods used to install the flange cleat on the column had the diameter of 25 mm. They were placed to go through the column and tightened with a wrench at both ends, using class 10 hex nuts only for specimens SX1 and SX2, as shown in Fig. 2a. Tensile coupon tests were conducted to measure the properties of the steel in accordance with the standard AS1391– 1991 [28] and the results are shown in Table 1. The specimens were designed to have a strong column weak beam system. Thanks to the concrete inside the column sections, the moment capacity of the columns is much larger than that of beams with a relative capacity ratio of 11.1 for specimens SX1 and SX2 and of 6.2 for specimens SY1 and SY2. Pretensions of the M20 and M22 bolts are determined in Eq. (2) according to Eurocode 3 [29], in which fub is the ultimate tensile strength of high-strength bolts (i.e.1000 MPa for this study); As is the tensile stress area in the threaded region of the bolt and can be determined according to ISO 898-1 [30], resulting in 245 mm2 for M20 bolts and 303 mm2 for M22 bolts; γM7 is a safety factor recommended as 1.1 [29]. In this way pretension force Fp,Cd was calculated as 156 kN for M20 bolts and 193 kN for M22 bolts and was applied to bolts by a torque wrench. F p;Cd ¼

0:7 f ub As γM7

Fig. 4. Specimens (a) SX1; (b) SY1; (c) SX2; and (d) SY2.

ð2Þ

C. Ding et al. / Journal of Constructional Steel Research 162 (2019) 105751 Table 1 Initial rotational stiffness of different specimens (unit: 106 N·m).

Steel plate for beam flange Steel plate for beam web Steel plate for column M20 & M22 bolts Threaded steel rods

Nominal yield stress (MPa)

Tested yield stress (MPa)

Tested ultimate stress (MPa)

345 345 345 990 600

359 351 357 N/A N/A

485 470 502 N/A N/A

3.2. Experimental setup The specimens were designed to be restrained at the two column ends and loaded at the two beam ends for SX1 and SY1, or at one beam end for SX2 and SY2. The experimental setup is shown in Fig. 5. To provide adequate restraints and reaction forces to the specimens, the frame system consists of four reaction frames in parallel (see Fig. 5a) that are anchored with the strong floor, several transverse beams (HB1, HB2, HB4, HB4-a, HB6) in between (to improve lateral stiffness) and base beams between different reaction frames (see Fig. 5a). Loading equipment includes a hydraulic jack to apply compression at a constant level on the column top end of the specimens, and two MTS actuators to apply vertical loads at the beam ends of the specimens. Each actuator was pinned to the beam end at one side and also pinned to the reaction frame at the other side. Both two column ends of the specimens were connected to the reaction frame, where the top end was restrained from horizontal movement but in-plane rotation was allowed as it was pinned to the two transverse beams (HB6s see Fig. 5b). The hydraulic jack was placed at the top end of the column with a relatively small contact area, and was therefore unable to provide effective restraint for rotational movement of the column top end. The bottom of the column was restrained from horizontal movement through the steel base plate bolted with the column end using two M20 bolts (see Fig. 5c) along the neutral axis of the column section. Again, this setup may not be effective to restrain rotational movement of the column bottom end.

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During the experiments, a compressive load of 1000 kN was applied first on the column top of the specimens, corresponding to an axial compression ratio of 0.4 (relative to the nominal strength) for the CFST columns, representing the gravity load received from upper floors. This load was kept constant during the experimental process. Subsequently, in specimens SX1 and SY1 with interior columns with 2 beams installed, vertical loads were introduced in opposite directions by the 2 actuators at the ends of the 2 beams, with identical loading rates of 1 mm/s. The experiments stopped when the specimens lost their load capacity or when any of the actuators travelled to its maximum distance. In specimens SX2 and SY2 with an exterior column with only one beam installed, the experimental setup was similar but only one actuator was used to apply one vertical load in the upward direction at the beam end. To ensure safety, the centre of the load point on the beam was 200 mm from the beam end. 3.3. Instrumentation Strain gauges, linear variable displacement transducers (LVDTs) and a total station digital theodolite (Hi-Target ATS-320R, with precision for angle measurement of 5.6 × 10−4 degree and for distance measurement of 2 mm) were used to measure the strains and displacements for specimens during experiments. The locations of strain gauges are shown in Fig. 6a, where gauges on the flange and web of the steel beams were used to evaluate the bending strain and potential yielding of beams as well as for understanding of the beam curvature. Other strain gauges near the bolt holes were used to examine whether bolts were in contact with bolt holes. A strain rosette was installed at the centre of the panel zone in the column to measure the shear behaviour of the panel zone. Two LVDTs were installed close to the beam flanges (Fig. 6b) to record the relative displacements between the beam and the column. Such results were necessary for further calculation of the relative rotation between beam and column. Loads and travel distances were automatically recorded through the corresponding actuators. Importantly, loading was paused every 30 s and the theodolite (see Fig. 5a) was used to manually verify the horizontal displacements of 2 column ends and vertical displacements of 2 beam ends, considering that the column ends might slip and move slightly due to the relaxation

Fig. 5. Experimental setup (a) overview; (b) connection detail of column top; and (c) column base connection.

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Fig. 6. Locations of (a) strain gauges; and (b) LVDTs.

of the horizontal restraints as the stiffnesses of the reaction frames cannot be infinitely large. Also, the travel distances recorded by actuators might not accurately represent the vertical displacements of beam ends due to possible slip between actuator and beam end resulting in a slight incline of the actuator during loading. After verification of the results recorded by the actuators, the vertical displacements of the beam ends were calibrated by a factor of 1.07 to make them consistent with the results measured by the theodolite. 4. Finite element (FE) modelling FE models were established for all specimens using ABAQUS 2017. Two examples are shown in Fig. 7 for specimens SX1 and SY1. Elements used in these FE models are eight-node solid elements with full integration (C3D8), eight-node solid elements with reduced integration (C3D8R), two-node beam elements (B31) and two-node spring elements (Spring2). It should be noted that a distance of 510 mm of each beam end measured from the surface of the column is modelled with C3D8R elements, to capture the formulation of plastic hinges on the beam and the potential local buckling. The remaining parts of the steel beams might perform more consistently as typical beams and therefore

they are modelled with B31 beam elements for computational efficiency. To achieve deformation compatibility between these two element types, the end cross-section of the solid elements is coupled with the node at the end of the B31 beam element (Fig. 7), making the rotational and translational movements of the end cross-section equal to those of the end node in the beam element. All the fin plates, flange cleats, side plates, steel rods, bolts, column steel tubes, concrete cores and column bases were modelled with C3D8R elements so that their stress distribution could be captured. Considering that internal stiffeners (see Fig. 7b) are key components for load transfer in specimens SY1 and SY2, they were modelled with C3D8 elements to achieve better computation accuracy. Threaded steel rods were modelled with B31 two-node beam elements because they were subject to axial forces and bending moments only, and the corresponding nuts were modelled with C3D8R elements. These two kinds of elements (B31 and C3D8R) were coupled again in the same way as for the B31 and C3D8R elements in the FE modelling of the steel beams. Each fillet weld was modelled using a series of spring elements with linear elastic material properties with a very large stiffness (represented in Fig. 7b and d by small dots). These spring elements bind together a series of node pairs between the two different members

Fig. 7. Finite element model (a) SY1 model overview; (b) detail of SY1; (c) SX1 model overview; and (d) detail of SX1.

C. Ding et al. / Journal of Constructional Steel Research 162 (2019) 105751

such as the flange cleat and the column steel tube as shown in Fig. 4a in order to understand the internal forces and stresses there and the associated failure mechanism. The stiffness value was defined to limit the maximum relative displacement between the node pairs in this study to be less than 0.05 mm, corresponding to 1/200 of weld leg length; a similar approach was used in [31]. Surface-to-surface contact interactions were adopted between all contact surfaces. The frictional coefficient of 0.3 [32] was used for steel-to-steel contact, such as that between the bolt head and the surface of the beam flange or between the bolt shank and bolt hole. The frictional coefficient for steel-to-concrete contact was defined as 0.57 [33] for the surface between the concrete core and steel tube, and the surface between the concrete core and internal stiffeners was treated as steel-to-concrete contact. As the surfaces of the steel threaded rods were smooth because of the zinc coatings, the frictional coefficient was considered to be zero. Pretension of bolts was applied according to Eq. (2) in Section 3.1 and was introduced by decreasing the temperature, as a common approach in such FE analysis [34]. Bilinear material constitutive models were used for the material properties of all steel, with the hardening ratio of 1% [33]. The constitutive model for confined concrete was determined according to the stress-strain relationship in [35], considering the concrete cracking. 5. Results and discussion 5.1. Load-displacement curves The mechanical behaviours of specimens can be described in terms of load-displacement curves as shown in Fig. 8, where the load is the vertical load applied to the beam end and the displacement is the

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corresponding vertical displacement. Therefore, the slopes of the curves represent the stiffness (loaded at beam ends) of the specimens. Overall, the load capacities of 4 specimens are similar (i.e. about 150 kN). As the stiffness of a beam-to-column specimen is contributed from the stiffnesses of its beam and column members as well as the connection between them, it can be found that the initial stiffness of specimens with internal stiffeners (i.e. 2.6 kN/mm for specimen SY1, 3.6 kN/mm for specimen SY2) is about 25% lower than that of specimens with threaded steel rods (i.e. 3.7 kN/mm for specimen SX1 and 4.6 kN/mm for specimen SX2). This suggests that the beam-to-column specimens with threaded steel rods might provide higher stiffness than the specimens with internal stiffeners. However, such a stiffness difference of specimens is contributed not only from the different connection methods (with rods or stiffeners) but also the difference in column stiffness. For specimens SX1 with steel rods in the connections, the column section provides its moment of inertia in the strong direction, being about four times of that in its weak direction for the specimens SY1 with internal stiffeners. Further discussion of the moment-rotation stiffness for the connections along is presented in Section 5.3, where the effects from column stiffness can be excluded. The load-displacement curve of specimen SX1 is linear elastic before the load reached 60 kN. When the loads increased to 60 kN and the corresponding displacements at the 2 beam ends reached 16 mm, fillet welds between the flange cleats and the column were subject to large horizontal internal forces and crack failure occurred, at which time stiffness began to decrease while the loading kept increasing. When the applied displacements at the beam ends reached 50 mm, the force dropped sharply due to an unexpected release of load in the actuators. That was not the intention in the experiment but it was corrected shortly nevertheless. When the displacements reached 110 mm, the cracks in the fillet welds between column and flange cleats became

Fig. 8. Load-displacement curves of specimens (a) SX1; (b) SX2; (c) SY1; and (d) SY2.

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corresponding experimental results. The ultimate capacity of SX2 was slightly (about 10%) lower than the FE modelling results, while the ultimate capacity of SY2 was comparable to that of SY1 and was slightly (about 10%) higher than the FE results. The ultimate capacity of SX1 was not reasonably described in the FE results because the fillet welds were modelled with spring elements with a linear elastic property. However, this modelling approach does indicate the high stress at the welding locations (therefore potential failures there) and does provide such insights to further understand the failure mechanism. 5.2. Failure modes and yielding mechanisms

Fig. 9. Failure modes of SX1 and SX2: crack of fillet weld.

significant. For safety considerations, therefore, the loading process was stopped at 120 mm, corresponding to a storey drift of 5.2% (the storey drift was calculated based on Eq. (1)). The load-displacement curve of specimen SX2 before the load reached 70 kN is linear elastic. When the displacement reached 30 mm, the load in the actuator suddenly released but was also recovered shortly (similar to that observed for SX1). Fillet welds between flange cleats and the column on SX2 did not experience any premature failure but finally cracked when the displacement reached 135 mm, corresponding to a storey drift of 5.8%. The experiment was then stopped. The load-displacement curve of SY1 is shown in Fig. 8c. After the applied displacement reached 135 mm, the maximum vertical load was reached (150 kN). Subsequently the loads began to drop, at which time plastic hinges were formed at the 2 beam ends. To ensure safety, the loading process was stopped at 180 mm, corresponding to a storey drift of 7.7%. The beam of SY2 (Fig. 8d) was eccentrically installed to the column, which therefore was subject to torque and experienced slight rotational movement along its longitudinal axis, causing the stiffness to decrease after the displacement reached 12 mm. A plastic hinge was formed at the beam when the displacement reached 120 mm, after which the load began to drop. The experiment was stopped when the actuator travelled to its maximum displacement of 200 mm, corresponding to a storey drift of 8.6%. All the specimens indicated satisfactory ductility, evidenced by their maximum storey drifts of more than 5% achieved in the experiments. The load-carrying capacities of specimens SY1 and SY2 were determined by the plastic hinges formed at the beam ends. The load-carrying capacities of specimens SX1 and SX2 were determined by the cracking of the fillet weld. FE results of load-displacement curves are also shown in Fig. 8. The initial stiffnesses of all the specimens (Fig. 8) are agree well with the corresponding experimental results. Meanwhile, the capacities of specimen SX2, SY1 and SY2 estimated from FE modelling are also close to the

As shown in Fig. 9, specimens with threaded steel rods (SX1 and SX2) failed by cracked fillet welds between the flange cleat and the column steel tube. Specimens with internal stiffeners (SY1 and SY2) did not fail during the tests, but plastic hinges formed at the beam end on the flanges subject to compression (Fig. 10a). To further examine whether there were any other failures in these 4 specimens, the specimens were disassembled after the experiment. It was found that all the bolt holes at different positions maintained their initial shapes (Fig. 11), and the bolt shanks did not break, indicating that the bolt holes and bolts did not fail. The results of FE analyses also showed that the shear stress in the fillet weld that finally cracked was large. As stated in Section 4, spring elements (Fig. 7b and d) were used to model the fillet welds. Shear stresses at two different loading levels (60 kN, 90 kN) were determined based on the horizontal internal forces in the spring elements modelling the fillet weld on the X axis in Fig. 12a (specimen SX1), and are plotted in Fig. 12b (tension is positive). For comparison, the shear stresses in the fillet welds on the X axis in specimen SY1 (Fig. 12c) are also plotted in the same figure. It can be seen in Fig. 12b that the shear stress distributed in the fillet weld in specimen SX1 (the specimen with the interior column using steel rods) is much more unevenly distributed than that in specimen SY1 (the specimen with the interior column using internal stiffeners), especially from X of 0 to 30 mm where the shear stress is very large on one side but small on the other. The fillet weld of specimen SX1 at X of 0 mm was subjected to the maximum shear stress and it is the position that began to crack during the experiment. At this point, when the load level reaches 60 kN, the maximum shear stress in the fillet weld of SX1 is 490 MPa, corresponding to 2.4 times the maximum shear stress in SY1 (202 MPa at X of 100 mm) and this difference increases when the load is increased to 90 kN. That may be the reason for the fillet weld cracking in specimen SX1 but not in SY1. The FE results of specimens SY1 and SY2 showed that plastic hinges were formed at the beam ends (Fig. 10), in consistence with the experimental results. Further comparison may indicate that the positions of the plastic hinges in the experiments were slightly different from that in the FE models. The experimental results showed the locations

b)

a)

Column

Column

Beam

Beam

a)

Column olumn

Local Buckling

Fig. 10. Comparison of failure modes for specimens SY1 and SY2 between (a) FE modelling; (b) experiment.

C. Ding et al. / Journal of Constructional Steel Research 162 (2019) 105751

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Fig. 11. Bolt holes after experiment (a) in the beam; (b) in the fin plate.

partially inside the bolted connection and partially outside the region stiffened by the bolted connection (Fig. 10a); while those in the FE models were outside the stiffened region (Fig. 10a). This may be because the bolt pretentions in the experiments were less than those in the FE models. Nevertheless, such a difference in the locations of beam end is still minor. Because the strain gauges installed on the beam web and upper beam flange (Fig. 6a) were along a cross section within the range of plastic hinge, their strain values at different load levels (40 kN, 60 kN, 80 kN and 100 kN) are plotted in Fig. 13 to understand the strain distributions at this region and the variations of the strain distributions at different load levels. The vertical axis refers to the positions of the strain gauges along the beam depth where the value of 0 corresponds to the neutral axis. Strain results from FE models are also plotted in the figure for a comparison with experimental results. At the load levels of 40 kN, 60 kN and 80 kN, it can be found that the strain distribution curves from specimen SY1 and SY2 are almost linear, indicating that the strain was proportional to the depth to the neutral axis and therefore no plastic hinge was formed at these load levels. However, at the load level of 100 kN, the strain distributions from both specimens show nonlinearity at the locations near the lower beam flange (i.e. 150 mm below the neutral axis, or −150 mm for the position in Fig. 13). This suggests that the lower flange of the beams in specimens SY1 and SY2 has yielded at this load level. When the load further increased, such local buckling became more obvious and finally a plastic hinge was formed there for specimens SY1 and SY2. 5.3. Moment-rotation stiffness According to Eurocode 3 [29], a connection can be classified as rigid if the initial rotational stiffness Sj,ini satisfies the condition: S j;ini ≥

kb EIb Lb

ð3Þ

Moreover, a connection can be classified as pinned if the initial rotational stiffness Sj,ini satisfies the condition: S j;ini ≤

0:5EIb Lb

ð4Þ

In Eqs. (3) and (4), E is Young's modulus of the steel beam, Ib is the moment of inertia of the beam, Lb is the span of the beam, equal to the centre-to-centre distance of 2 adjacent columns. kb is a factor related to the type of the structure, as 25 for a frame without bracing system. The initial rotational stiffness Sj,ini for specimens was determined as the slope of the moment-rotation curve at the elastic stage, where the rotation and moment could be calculated using the measured results of LVDTs from Fig. 6b through the equations: θc ¼

d1 −d2 l1;2

M ¼ F v Lbeam

ð5Þ ð6Þ

where d1 and d2 are the displacements recorded by LVDTs shown in Fig. 6b, l1,2 is the distance between 2 LVDTs, Fv is the vertical load applied by the actuator and Lbeam is the length of the beam. Table 2 shows the resulting initial stiffness of each specimen compared to the values of rigid and pinned conditions according to the calculation results of the initial rotational stiffness. In the Y direction, where internal stiffeners were used to form the connection, the initial rotational stiffness of each specimen (SY1 with the interior column or SY2 with the corner column) satisfied the rigid condition. Therefore, such configurations provided high momentrotation stiffness and these 2 connections could be classified as rigid for interior and exterior applications. In the X direction where threaded steel rods were installed, although specimens SX1 and SX2 should be classified as having semi-rigid connections, their initial rotational

Fig. 12. (a) Position of fillet weld in SX1; (b) coupling forces of spring elements for modelling fillet weld; and (c) position of fillet weld in SY1.

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C. Ding et al. / Journal of Constructional Steel Research 162 (2019) 105751

Fig. 13. Strain distributions at beam end in the same cross section at different height of (a) SY1; and (b) SY2.

stiffness was close to the rigid boundary (about 85% rigid condition). Adequate moment-rotation stiffness provided by the steel rod configurations was still well demonstrated. 5.4. Load transfer of threaded steel rods and internal stiffeners Threaded steel rods and internal stiffeners were embedded in the concrete and were important components in the connection configurations for both directions, as they could compensate for core concrete in tension and prevent excessive deformation of the steel tube flanges subjected to the forces transferred from the beam ends. In this section, internal forces, including bending moments and axial forces, of these components are clarified and compared based on the FE results. The axial force and bending moment of each component were calculated, based on the stress field for three loading levels (60 kN, 90 kN, 120 kN at the beam end) as shown in Figs. 14 and 15. In the figures, the upper stiffener is the internal stiffener above the beam and the lower stiffener is the internal stiffener below the beam for specimens SY1 or SY2. Also, because specimens SX1 and SX2 were stiffened by 2 pairs of steel rods (2 upper rods and 2 lower rods, see Fig. 2a), the axial force or bending moment of the steel rods shown in Figs. 14 and 15 is the sum of the 2 rods in the same horizontal row. When the load was increased to 120 kN in the experiments, the bending moment at the beam end became 253.2 kN·m for specimens SX1 and SX2, i.e. 120 kN × (2.53 m–0.2 m-0.22 m); and 266.4 kN·m for specimens SY1 and SY2, i.e. 120 kN × (2.53 m–0.2 m-0.11 m). Considering that the beam flanges only carried bending moment whereas the beam web only carried shear force [36], the bending moments at the beam end were formed by a compression force at one beam flange and a tension force at the other. The compression force or tension force was then 723 kN for specimen SX1 and SX2, i.e. 253.2 kN·m/ 0.35 m, and 761 kN for specimens SY1 and SY2, i.e. and 266.4 kN·m/ 0.35 m. Such levels of compression force and tension force were finally transmitted from the beam flanges to the column. The compression force carried by the column flange and the concrete inside the column may not be critical for the connection. However, the tension force in specimens SX1 or SX2 (denoted as FX,B in Fig. 16a) need to be carried by both the column flange and the steel rods, and the tension force in specimens SY1 or SY2 (denoted as FY,B in Fig. 16b) need to be carried by both the column flange and the internal stiffeners. The column flanges may bend outward due to the large tension on the column flange (shown as distributed force qX, C or qY, C in Fig. 16), as the inside concrete cannot aid in resisting tension. Therefore, it is desirable to transfer more tension to the rods or stiffeners than to the column flanges. As indicated in Fig. 14, it was found that about 30% of tension (238 of 723 kN) was carried through the steel rods for specimen SX1 and 65% of the tension (490 of 761 kN) through the internal stiffeners for specimen SY1. Similarly, 17% of the tension (140 of 723 kN) was transferred to the

steel rods in SX2 and 65% of the tension (490 of 761 kN) was carried by the internal stiffeners in SY2. These results clearly indicate that the internal stiffeners within the CFST could carry efficiently and transfer the majority of the tensile force (over 60%), and this configuration is clearly more effective than that using steel threaded rods. It should also be noted that the column flanges (220 mm) in the X direction (Fig. 2) are much narrower than those in the Y direction (440 mm). Therefore, the column flange in the X direction may be more difficult to bend outward (due to its smaller span between the 2 webs). This may also explain why a higher proportion of tension was carried by the column flanges for specimens SX1 and SX2 with the connection in the X direction. Each of the vertical y axes in Fig. 15b to e show the bending moments carried by the section of steel rods or internal stiffener along their longitudinal direction, i.e. x coordinates shown in Fig. 15a. It can be found that the steel rods in specimen SX1 or SX2 always carry much less bending moment than the stiffeners in specimen SY1 or SY2, because a pair of rods offers a much lower moment of inertia than a stiffener, and the nuts tightening the steel rod might not effectively restrain the rotational movements of the two sides of the rod. Therefore, the fillet welds shown in Fig. 12a in specimen SX1 or SX2 had to help the rods carry more bending moment, resulting in an unevenly distributed horizontal force in the fillet weld as shown in Fig. 12b. It can be also found in Fig. 15b that the bending moment on the left side of the lower stiffener in specimen SY1 is about five times that on the right side. This is related to the force to which the column steel tube and the core concrete were subject. As the left actuator applied the vertical load upward while the right actuator applied downward (Fig. 5a), thelower flange of the right beam was under compression while the lower flange of the left beam was under tension. The right side of the lower stiffener carried bending moment with the assistance of the surrounding concrete that was under compression, whereas the left side of the lower stiffener could only carry bending moment by itself because the surrounding concrete was under tension and might already have reached its tensile limit. However, the left side of the upper stiffener carries bending moment with the assistance of the surrounding concrete that is under compression, whereas the right side of the upper stiffener can only carry bending moment by itself. Therefore in Fig. 15c the bending moment on the right side of the upper stiffener in specimen SY1 is much greater than that on the leftside. Table 2 Initial rotational stiffness of different specimens (unit: 106 N·m).

Nominally pinned condition Rigid condition Initial rotation stiffness: Sj,ini

SX1

SX2

SY1

SY2

6 282 238

6 282 241

5 268 1051

5 268 1150

C. Ding et al. / Journal of Constructional Steel Research 162 (2019) 105751

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Fig. 14. Axial forces in stiffening members.

The beam end of specimen SX2 with a corner column and steel rods and the beam end of specimen SY2 with a corner column and internal stiffeners were loaded upward in the experiments. The upper flange of the beam carried compression force while the lower flange of the beam carried tension force. Comparison of

Fig. 15d and e showed that the lower stiffener carried a greater bending moment, about double that carried by the upper stiffener. This was also caused by the lower stiffener which carried bending moment with the assistance of the surrounding concrete which was under compression, while the upper stiffener only carried

Fig. 15. Bending moments of (a) longitudinal direction of stiffener and rod; (b) lower stiffener and rods in SX1 and SY1; (c) upper stiffener and rods in SX1 and SY1; (d) lower stiffener and rods in SX2 and SY2; and (e) upper stiffener and rods in SX2 and SY2.

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C. Ding et al. / Journal of Constructional Steel Research 162 (2019) 105751

Fig. 16. Free body diagrams for (a) flange cleat of specimen SX1 or SX2; and (b) side plate and internal stiffener of specimen SY1 and SY2.

bending moment by itself as the surrounding concrete was under tension. 6. Conclusions The present study proposed and examined a new prefabricated connection system between steel beams and a concrete-filled steel tube (CFST) column. In this new connection system, the panel zone is stiffened by threaded steel rods in one horizontal direction and by internal stiffeners in the other horizontal direction. Additional steel plates including flange cleats, side plates and fin plates are fillet welded to the CFST column in the factory and steel beams in two directions can be joined to the column mainly through bolting. To study momentrotation behaviours of the connection in two directions, 4 specimens were prepared considering both interior connection and exterior connection scenarios. FE modelling was further performed to understand the performance of the connection system and the internal forces in the fillet welds, internal stiffeners and steel rods in particular. On the basis of the experimental and modelling results, the following conclusions can be drawn: 1) The proposed connection system showed high moment-rotation stiffness in both horizontal directions. According to the classification of connection rigidity in Eurocode 3, if this connection system is used in moment-resisting frames, it can achieve rigid connection in the same direction as the configuration of internal stiffeners; and it can be a semi-rigid connection with 85% of the stiffness of the rigid condition in the same direction as the configuration of threaded steel rods. 2) Internal stiffeners and threaded steel rods in the connection can effectively stiffen the panel zone, and none of the column flanges in the specimens experienced noticeable bending deformation. This is because either the internal stiffener or the steel rods, rather than the column flanges, could help carry tension force transferred from the beam flange. In specimens SY1 and SY2 in the direction stiffened by internal stiffeners, about 65% of the tension force from the beam flange was carried to the internal stiffeners and the other 35% was carried to the column flanges through fillet welds. In contrast, in specimens SX1 and SX2 stiffened by steel rods, the proportion of tension transferred from the beam flange to the steel rods was much lower, being less than 30%, while the remaining 70% of the tension was transferred to the column flanges. 3) Internal stiffeners in the connection system can effectively transfer both axial force and bending moment, whereas threaded steel rods may mainly transfer axial force. This is because a pair of rods offers

a much lower moment of inertia than a stiffener, and the nuts tightening the steel rod might not effectively restrain rotational movements of the two sides of the rod. Therefore, the vertical fillet weld between the flange cleat and the column steel tube in specimen SX1 or SX2 needs to assist the threaded steel rods to carry more bending moment, resulting in unevenly distributed horizontal force in the vertical fillet weld plus associated cracking. 4) The proposed connection system also showed satisfactory ductility for the moment-rotation response in the direction stiffened by internal stiffeners. In this study the connection system in this direction was able to develop a storey drift of 7% without structural failure. The final failure was due to local buckling of the beam flanges where a plastic hinge was formed about 350 mm from the surface of the column in a satisfactorily ductile manner. In the direction stiffened by threaded steel rods, the corresponding story drift when the vertical fillet welds failed could still reach 5% and this also provided a high deformation capacity. To achieve convenient assembly of the connection system and its acceptable mechanical performances, a few design considerations may be expressed. For example, it may be better to have slotted holes in the upper flange cleats for the threaded steel rods rather than circular holes, as such holes allow the threaded steel rods to easily pass through the flange cleats and column flanges and provide tolerances to accommodate possible fabrication inaccuracy. Moreover, the fillet welds between the flange cleats and column flanges require good quality because of high stresses there. In addition, a thicker flange cleat with increased bending stiffness may introduce better uniformly distributed stress in the fillet welds, mitigating high stress concentration at the fillet weld locations. Acknowledgement The authors are grateful for support from the Australian Research Council through the Discovery Project (DP180102208). Thanks are also given to the technical support for the experiments at the Faculty of Civil Engineering in Heifei University of Technology and to Anhui Fuhuang Steel Structure Co., Ltd. for manufacturing experimental specimens. References [1] S.P. Schneider, Axially loaded concrete-filled steel tubes, J. Struct. Eng. 124 (10) (1998) 1125–1138. [2] F. Ding, Z. Yu, Y. Bai, Y. Gong, Elasto-plastic analysis of circular concrete-filled steel tube stub columns, J. Constr. Steel Res. 67 (10) (2011) 1567–1577. [3] T. Sheehan, X. Dai, T.M. Chan, D. Lam, Structural response of concrete-filled elliptical steel hollow sections under eccentric compression, Eng. Struct. 45 (2012) 314–323.

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