Seismic performance of a FRP encased concrete bridge pylon connection

Composites: Part B 38 (2007) 685–702 www.elsevier.com/locate/compositesb

Seismic performance of a FRP encased concrete bridge pylon connection Yael Van Den Einde 1, Vistasp M. Karbhari *, Frieder Seible Department of Structural Engineering, University of California, San Diego, La Jolla, CA 92093-0085, USA Received 10 March 2006; accepted 2 July 2006 Available online 26 January 2007

Abstract Fiber reinforced polymer (FRP) composites provide immense advantages in the development of bridge structural components. In addition to their attractiveness due to light weight and tailorable performance attributes, they also provide significant advantages for use in seismic regions. This paper describes the results of tests conducted to validate their use as pylons for cable-stayed bridges. A full-scale test of one segment of the pylon was conducted to evaluate the ductile performance of the splice connection as a potential region for inelastic action. Results indicate that all requirements for a successful proof of concept test were met. The force capacity was significantly greater than the maximum force demand for the bridge and the displacement capacity was close to six times the displacement demand under required seismic levels. Furthermore, the strain levels in the composite shell were well below the materials allowables set based on damage tolerance requirements. Ó 2007 Elsevier Ltd. All rights reserved. Keywords: A. Polymer-matrix composites (PMCs); A. Hybrid; B. Strength; C. Analytical modelling; Pylon

1. Introduction Fiber reinforced polymer (FRP) composites have been shown to provide structurally-efficient alternatives to conventional construction materials as related to bridge decks, rehabilitation of concrete decks and girders and the seismic retrofit of columns. In these applications the attributes of low weight, tailorable performance characteristics based on designed anisotropy, corrosion resistance and light weight are significant advantages. While there is no doubt that the area of rehabilitation using externally bonded composites is the most significant growth area today, there is immense potential for the development of FRP composites as stay-in-place structural formwork incorporating the dual functions of formwork and reinforcement. This not *

Corresponding author. Tel.: +1 858 534 6470; fax: +1 858 534 6373. E-mail address: [email protected] (V.M. Karbhari). 1 Now with the NEES Cyberinfrastructure Center, San Diego Super Computer Center, University of California, San Diego, 9500 Gilman Drive, La Jolla, CA 92093-0505, USA. 1359-8368/$ - see front matter Ó 2007 Elsevier Ltd. All rights reserved. doi:10.1016/j.compositesb.2006.07.010

only builds on the inherent advantages of FRP composites but also provides for optimized integrated usage of FRP composites and concrete, in their most efficient modes of tension and compression, respectively. The combined use of FRP composites and concrete has specific relevance as an analog to the concept of concretefilled steel tubes which have been extensively researched over the past 3 decades. While the steel encased concrete concept is an attractive one for new construction it has not gained significant popularity due to concerns related to weight, corrosion and lower efficiency in confinement at low loads due to the higher Poisson’s ratio of steel as compared to that of concrete. The use of FRP composites, however, negates most of these concerns and in addition makes possible the incorporation of newer and perhaps more aesthetic designs due to the greater ease of fabricating special shapes. In addition, with FRP composites it is possible to tailor the hoop and longitudinal moduli (as compared to isotropy in steel) enabling better seismic performance. In recent years there has been an increased focus on the building of long-span bridges as a means of spanning large

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List of symbols Ag EH EL fc0 Fy F 0y P SBSL SBST Vf

gross area of concrete transverse or hoop modulus longitudinal modulus unconfined concrete strength predicted ideal yield force from a moment– curvature analysis target first yield force from a moment–curvature analysis axial load longitudinal short beam shear strength transverse short beam shear strength volume fraction of fiber

bodies of water, limiting the footprint of bridge structures on the area spanned, and increasing seismic safety of the overall bridge system. Long-spans are increasingly being considered as the domain of cable-stayed bridges [1] with recent bridges such as the Cooper River Bridge in Charleston, South Carolina and the Rion-Antirion Bridge in Greece setting new standards both for aesthetics and functionality. In these bridges the cables are attached to towers which intrinsically bear the load with tension in the deck cables producing compression in the deck and towers. The towers are thus the key element of the bridge and often take a year or longer to raise. FRP composites have significant potential for use in this type of component serving as both the stay-in-place structural form and as the reinforcement. The lighter weight would make construction easier and the ability to tailor anisotropy makes it possible to optimize each pylon section for the specific load demands in that local region.

ideal yield or ductility 1 displacement Dy D0y;push experimental first yield displacement in the push direction D0y;pull experimental first yield displacement in the pull direction ec compression strain et tension strain l/ curvature ductility lD displacement ductility / curvature q density of composite

This paper reports on tests conducted on a pylon segment designed for a hypothetical FRP cable stayed bridge to span a major freeway. The bridge, shown schematically in Figs. 1 and 2 represents a 137 m (450 ft) long dual plane, asymmetric cable-stayed bridge supported by a 59 m (193 ft) high A-frame pylon, utilizing fiber reinforced polymer (FRP) composite materials. The cable stays are anchored to the steel pylon head that is connected to the concrete filled carbon/epoxy tubes forming the pylon legs. The pylon is approximately 59 m (193 ft) tall and the pylon legs at the base span a distance of 35.8 m (117 ft). The pylon leg, composed of 9.8 m (32 ft) sections that are spliced together using internal reinforcement, consists of circular filament wound carbon/epoxy shells, which are filled with concrete on-site. The carbon/epoxy shell has internal circumferential ribs to provide good mechanical interlock between the concrete and the carbon/epoxy shell, and the orientation of the carbon fibers in the shell is

Fig. 1. Schematic of bridge elevation.

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Fig. 2. Schematic of bridge cross-section showing A-frame pylon.

designed to provide both longitudinal reinforcement for flexure and transverse or hoop reinforcement for confinement of the infill concrete. Thus the carbon/epoxy shell serves both as a stay-in-place formwork for the concrete and the reinforcement, while the infill concrete serves to carry compressive forces and to stabilize the thin carbon/ epoxy shell against local buckling. The bridge has to be designed for specified limit states such as Service, Strength, Extreme, and Fatigue to achieve the objectives of constructability, safety, and serviceability with regard to issues of inspectability, economy, and aes-

thetics. The FRP composite components, being largely elastic to failure, must be designed to stay at strain levels below those likely to cause irreversible damage. Ductility of the system is provided by the inelastic design of the connections that uses mild steel reinforcement following AASHTO LRFD specifications [2]. It is noted that the splice connection serves as a location for possible inelastic action and is designed such that the ultimate capacity of the connection section satisfies the maximum demand, and that the tensile and compressive strain levels in the shell section are below the allowable strain levels limited

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to a certain percentage of ultimate, which for carbon fibers is conservatively assumed as 0.01 in tension and 0.006 in compression. The pylon connection regions are designed to remain essentially elastic under earthquake loads (i.e., curvature ductility, l/ 6 1.5). Thus in addition to being a proof test for the concept, the main objectives of the Pylon Connection Test were to validate the design and analysis methods used to develop the connection splice reinforcement details, to examine the ductile failure of the mild steel reinforced splice section, to assess the confinement effects of the carbon/epoxy shell, and to evaluate strain levels in the carbon/epoxy shell. 2. Description of test system The A-frame pylon in the proposed bridge system (Figs. 1 and 2) is composed of 9.75 m (32 ft) long sections that are spliced together up the pylon height. Over the first pylon segment near the footing, longitudinal reinforcement is provided up the entire height of the segment to ensure that the capacity of the pylon leg is sufficient to carry all loads in the event of fire damage or other extreme accidental damage to the carbon/epoxy composite shell. The splice detail at the footing also includes an increased amount of reinforcement due to the starter bars from the footing. Therefore, this footing splice detail is significantly different from the remaining splices up the pylon height. Analyses show that under most load cases, and specifically the Extreme (Seismic) load case, the maximum moment demand on the pylon occurs at the pylon base where the additional confinement and reinforcement is provided (see Fig. 3). The second highest moment demand occurs up the height of the column near one of the ‘‘typical’’ splice regions. For this reason, the ‘‘typical’’ splice was evaluated in the Pylon Connection Test. Preliminary analyses the pylon system indicated an axial load of 17,195 kN

(3864 kips) with more refined analyses indicating loads of 18,303 kN (4104 kips) and 21,198 kN (4764 kips) without and with consideration of the weight of added permanent loads such as sidewalks, barriers, and utilities at the end of construction, respectively. Several factors influence the design of the connection splice detail. First of all, it is necessary to ensure that pullout of the splice longitudinal bars is avoided. This can be accomplished by a combination of appropriate lap splice lengths and provision of full confinement of the bars. From a conservative viewpoint a splice length of 1.8 m (6 ft), following recommendations in the AASHTO guidelines [3], Priestley et al. [4] and Burguen˜o [5], is used and additional hoops consisting of 13 mm (#4) at 120 mm (4.75 in.) spacing are provided around the lap splice bars to assist the shell in confinement. It is, however, noted that the carbon/epoxy shell alone is capable of contributing most of the confinement, and that any additional transverse steel reinforcement only adds to conservatism and ensures that enough confinement is provided for ductile behavior in the plastic hinge region. Since the bond for lap splice is often lost between strains of 0.001–0.002 [4], a dilation strain of 0.00125 is assumed in the analysis. The final design details for the ‘‘typical’’ splice connection, shown in Fig. 4, result in the requirement of thirty 29 mm (1.125 in.) diameter longitudinal bars confined by 13 mm (0.5 in.) hoops spaced at 120 mm (4.75 in.). As noted earlier there is a high degree of conservatism in the design against lap-splice failure due to the combined effect of the FRP composite shell and the additional confinement provided by the hoops. In order to facilitate construction of the bridge, a conventional longitudinal reinforcing cage (designed as ten 29 mm (1.125 in.) bars for the Pylon Connection Test) was considered for use over the full-height of each pylon segment. These bars are only provided in the pylon for ease 30

Approximate splice locations -9461

5664

25 20

Height up Pylon (m)

-10473

Test designed for this splice level

15

11816

10 -6640

9509

5 0

-20000

-15000

-10000

-5000 -4309

0 -5

5000

10000

15000

4470

-10

-17092

-15

11266

-20

Moment (kN-m)

Fig. 3. Maximum and minimum seismic moment distribution as a function of distance up the pylon height.

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approx. 1543mm [5'-043"] 9mm [0.354"] thick carbon/epoxy pylon 1524mm [5']

25.4mm [1"] clear cover from inside of shell to outside of hoops 30 29mm [#9] starter bars Staggered termination starting at 6' (1.83m) from top of footing

13mm [#4] hoops @ 120mm [4.75"]

Fig. 4. Schematic of design details at critical pylon cross-section.

of construction with respect to placement of the splice reinforcing steel cages between pylon segments. The bars are not relied upon to carry the necessary loading. Due to laboratory clearance issues, the full-height longitudinal bars for the Pylon Connection Test were tied, inserted in the pylon, and placed vertically simultaneously. The nominal longitudinal cage was essentially spliced over the full lap splice length in the connection region of 1.8 m (6 ft). Since the demand in the top and central region of the pylon is significantly lower than at the connection region and it is assumed that the carbon/epoxy shell provides enough confinement, a reduced amount of hoops was used for these continuous bars outside of the splice regions (13 mm (0.5 in.) diameter hoops at 452 mm (17.8 in.)) primarily to hold the continuous longitudinal bar cage together. The reinforcing details are shown in Fig. 5. The axial load was applied in the test using four external 27–15 mm (0.6 in.) strand Dywidag Systems International (DSI) tendons that were looped in the footing as shown in Fig. 5. Because the system of applying the axial load was passive, the exact axial load ratio applied on the specimen was unknown at the time the tendons were stressed due to potential losses in the anchorage and over the tendons. Post-tensioning was provided in both horizontal directions in the footing to help transfer the forces expected from the looped tendons. A concrete block load stub that was also post-tensioned in each horizontal direction was provided on top of the pylon as a reaction block for the external DSI anchorage systems, as well as a method to attach the hydraulic actuator for lateral loading. In addition to the post-tensioning, headed steel reinforcing bars were also provided in both the footing and load stub to help transfer forces and allow for shorter development lengths, which assisted in overcoming space constraints in the intricate footing. The axial load, which was applied using four external tendons with 27–15 mm (0.6 in.) diameter strands that were stressed individually to 80% of their ultimate strength, was approximately 19,856 kN (4460 kips) assuming a 10% loss during stressing and anchoring. The weight of the load stub was 185 kN (41.5 kips) and the full pylon weight was 419 kN (94.2 kips). This resulted in a total axial load of 20,452 kN (4596 kips) corresponding to an approximate

axial load ratio of P =fc0 Ag ¼ 0:27 assuming a design concrete strength of fc0 ¼ 41:4 MPa ð6 ksiÞ. The test setup for the Pylon Connection Test is shown in Fig. 6. In order to maintain a 38 mm (1.5 in.) gap at both the footing and load stub interfaces with the pylon segment, a neoprene rubber gasket was used as a spacer during grouting of the concrete. Strain gages were placed up the height of the pylon internally on the longitudinal reinforcement to measure uniaxial tensile and compressive strains and indicate where yielding occurred. Additionally, strain gages were applied to the transverse reinforcement and on the inner surface of the pylon in the transverse direction to measure core dilation and transverse strains. Longitudinal and transverse strain gages were placed on the outside of the FRP pylon in similar locations as the internal steel reinforcement gages. Four strain gages were also placed on each axial load tendon at a location of 3.7 m (146 in.) above the footing to assist in measuring the strains and consequently the forces in the tendon. As shown in Fig. 6 string potentiometers were connected at several locations up the pylon height relative to the reference wall to measure lateral displacements. Additionally, inclinometers were installed to measure rotations. To evaluate any overturning effects or uplift of the pylon footing, two vertical linear potentiometers and one horizontal linear potentiometer were placed at the footing-strong floor interface. Finally, vertical string potentiometers were positioned on either side of the pylon from the footing to the underside of the load stub to measure pylon elongation. The test unit was loaded quasi-statically with an incrementally increasing, fully reversed cyclic loading history as shown in Fig. 7. The lateral load was applied by a 2000 kN (445 kip) MTS 60.6 m (24 in.) stroke actuator that was positioned in the lower half of the load stub resulting in a 10.3 m (33.75 ft) length between the point of load application and the top of the footing. The initial three cycles were run in load control up to the theoretical first yield of the extreme longitudinal reinforcing bars at the critical section. The remainder of the test was conducted in displacement control. Single cycles were conducted at the Service (395 kN, 89 kips), Strength (622 kN, 140 kips), and Extreme (1222 kN, 275 kips) demand levels, as well as

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2.6 m [8'-6"]

1.52 m [5']

13 mm [#4] spiral @ nominal as fits in load stub

1.83 m [6']

38 mm [112"] gap

Ten 29 mm [#9] straight bars over full pylon height spliced top and bottom with 1.83 m [6'] connection starter bars

13.1 m [43'] full specimen height

15 mm [27-0.6"] φ loop tendon

Nominal 13 mm [#4] hoops at 452 mm [17.8"]

11.6 m [38'] from floor to bottom of load stub

9.75 m [32']

Thirty 29 mm [#9] headed starter bars terminating at 1.83 m [6'] up from footing

13 mm [#4] hoop @ 120 mm [4.75"]

27-15.2 mm [0.6"] φ loop tendon

1.83 m [6']

305 mm [1']

38 mm [112"] gap 1.83 m [6'] 13 mm [12"] layer of hydrostone 3.05 m [10'] Fig. 5. Schematic showing reinforcement details.

at first yield. After the first yield cycle, the experimental ideal yield displacement was calculated as 73 mm (2.9 in.) using Dy ¼

0 0 F y ðDy;push þ Dy;pull Þ 2 F 0y

ð1Þ

where Dy, D0y;push , and D0y;pull are the ideal yield or ductility 1 displacement, experimental first yield displacement in the push direction, and experimental first yield displacement in the pull direction, respectively and Fy and F 0y are the predicted ideal yield force and target first yield force from a moment–curvature analysis, respectively. For the purposes

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One 127 mm [5"] stroke string potentiometer from side wall to load stub to measure out of plane displacement Inclinometer @ full specimen height

1.83 m [6'] 1067 mm [3'-6"]

691

5.5 m [18'] MTS Actuator 2000 kN [450k]

Actuator Fixture

Two 1.02 m [40"] stroke string potentiomenter @ full height measured from top of footing

1.52 m [5']

4.27 m [14']

38 mm [112"] gap

1.22 m [4']

Inclinometer @ full specimen height Inclinometer @ 43 specimen height

7.73 m [25'-343"]

0.76 m [30"] stroke string potentiomenter @ 43 specimen height Two 127 mm [5"] stroke string potentiometers to measure elongation

13.1 m [43'] Specimen height 10.3 m [33'-9"]

Inclinometer @ 12 specimen height

5.15 m [16'-1012"]

0.64 m [25"] stroke string potentiomenter @ 12 specimen height

9.75 m [32'] 12.1 m [39'-9"] to centerline of actuator

15.2 m [50']

Inclinometer @ 14 specimen height

2.58 m [8'-514"]

0.38 m [15"] stroke string potentiomenter @ 14 specimen height Inclinometer @ 81 specimen height

1.29 m [4'-285"]

4 Curvature measuring devices (linear potentiometers 100 mm [4"] stroke)

38 mm [112"] gap 1.83 m [6']

Two vertical devices (on both sides of footing) and 1 horizontal device (linear potentiometers 100 mm [4"] stroke)

3.96 m [13'] Fig. 6. Schematic of test set-up with instrumentation.

of this the ‘‘push’’ side of the pylon refers to the side that is in compression when the actuator is pushing on the specimen and conversely, the ‘‘pull’’ side refers to the side of the pylon that is in compression during the pull response of the actuator.

The specimen was cycled three times at the displacement ductility 1 level, and two times for each subsequent ductility level up to displacement ductility 4. At this point the displacement capacity was already three times the displacement expected under the Extreme (seismic) demand of the

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10

Displacement Ductility μΔ

PUSH

Load control (3 cycles)

8

Extreme Demand

6 4 2

F'y Service Demand

μΔ=1.5 μΔ=2

μΔ=1

μΔ=3

μΔ=4

μΔ=6 μΔ=5

μΔ=7

0 -2 Strength Demand

-4 -6

μΔ=Δ/Δy

-8

PULL

-10 0

2

4

6

8

10

Displacement control (16 cycles)

12

14

16

18

20

Number of Cycles Fig. 7. Schematic of loading history.

bridge and therefore, single cycles were conducted for the remaining displacement ductility cycles to assuage concerns related to the stability of the axial load system. During loading to displacement ductility 7 in the pull direction, five of the longitudinal bars in the connection region were noted to rupture resulting in a slight decrease in overall capacity. To investigate the effect of this failure mechanism in terms of degradation in strength, an additional cycle was conducted at displacement ductility 7. 3. Details of materials and FRP component Mild steel reinforcement (A 706) was used for the main reinforcing steel. The yield and ultimate strengths, and yield, strain hardening, and ultimate strains for the reinforcing bars, as determined through sample testing are shown in Table 1. It is noted that the yield strength of the transverse reinforcement cannot be accurately determined because the steel reaches yield when shaped into a spiral and is re-yielded to straighten it into the coupons for testing. Material characteristics for the headed bars are also reported in Table 1.

The uncoated seven-wire steel strands used for the posttensioned tendons were provided by Dywidag Systems International (DSI). The strands were 15 mm (0.6 in.) in diameter and were made of ASTM A416 Grade 270 (1862 MPa, 270 ksi) steel. The cross-sectional area of each strand was 141 mm2 (0.2186 in.2) and the elastic modulus was 194 GPa (28.1 msi). The steel pipes in the footing were grouted with Riverside Type1/Type2 water cement grout having a design strength of 21 MPa (3 ksi). The pylon shell was filament wound on a steel mandrel using a specially designed elastomeric surface layer to enable formation of the internal ribs. The 21 layer layup was designed with a (892/±10/892/±10/892/±10/892/±10/ 895) layup which optimized both hoop constraint and longitudinal stiffness. It is noted that the average thickness of the 89° layers was 0.4175 mm (0.0164 in.) and the average thickness of the ±10° layers was 1.35 mm (0.053 in.), giving a total thickness of the pylon shell of 8.1 mm (0.319 in.). Characteristics as determined through use of classical lamination theory, tests on tag ends, and tests using a specially designed 508 mm ring-burst test are reported in Table 2. Figs. 8a and b show the pylon section being lifted for placement and being brought in alignment with the footing steel, respectively. An overall view of the specimen prior to test initiation is shown in Fig. 9. The design concrete strength was fc0 ¼ 34 MPa ð5 ksiÞ for the pylon and footing and fc0 ¼ 41 MPa ð6 ksiÞ for the load stub since a short cure time was expected and a concrete strength of at least fc0 ¼ 34 MPa ð5 ksiÞ was required to stress the axial load tendons. The concrete for the Pylon Connection Test specimen was poured in four stages, with the first stage being for the footing, the second up the pylon to a height of 4.88 m, the third being the remaining 4.57 m of the pylon, and the fourth being the gap between the pylon and the load-stub and the entire load-stub. Compression strength characteristics were assessed through tests on standard 152 mm (6 in.) diameter by 305 mm (12 in.) high cylinders with results as listed in Table 3.

Table 1 Material properties of reinforcing steel Bar

Size mm (#)

fy MPa (ksi)

fu MPa (ksi)

ey

esh

esu

Transverse hoops Longitudinal headed bars Longitudinal straight bars

12.7 (#4) 28.6 (#9) 28.6 (#9)

– 469 (68.3) 490 (71.1)

742 (107.6) 655 (95.0) 688 (99.8)

– 0.00265 0.0027

– 0.014 0.0099

0.1 0.1 0.1

Table 2 FRP pylon material properties

Thickness, mm (in.) Vf q, g/cm3 (lb/ft3) EL, GPa (msi) EH, GPa (msi) SBSL, MPa (ksi) SBSH, MPa (ksi) a

Analytical

Experimental based on burst test

Experimental based on tag ends

8.1 (0.319) 0.55 1.60 (100.0) 50 (7.24) 90 (13.03) NA NA

9 (0.35) 0.661 1.55 (96.7) 62 (8.98) 82 (11.9) 23 (3.34) 26.6 (3.86)

9 (0.35) 0.579 1.47 (91.7) 52.4 (7.6) 90 (13.03)a 20.1 (2.92) 26.6 (3.86)

Value was calculated using classical lamination theory.

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Fig. 9. Overall view of test specimen prior to loading.

Fig. 8. Construction of the pylon: (a) lifting the FRP shell; (b) placing the FRP shell over the reinforcing cage extending from the footing.

4. Analytical development 4.1. Moment–curvature analyses The analysis of the sections was conducted using moment–curvature programs developed specifically for FRP confined concrete shells. Throughout the analysis it is assumed that plane sections remain plane in bending. Also, after cracking in the concrete initiates, the concrete tensile strength is neglected. It is also assumed that slip

does not occur between the carbon/epoxy shell and the concrete so that full composite action is developed. This is ensured by the design of the internal ribs and was earlier confirmed in [6] and was further confirmed by removal of sections of the pylon shell after completion of the current test. Furthermore, the stress–strain behavior of the carbon/epoxy shell is found by using equivalent orthotropic plate properties for the shell, while the stress–strain behavior of the concrete is obtained by assuming the concrete is an isotropic cylinder. Concrete confinement is described by Mander’s incremental passive confinement model [7]. The flexural strength of the section is obtained at the point where the ultimate compressive or tensile strain of the shell is developed, and stiffening effects due to tension are ignored. To compute the moment–curvature response the curvature is evaluated at incremental values of compressive strain in the carbon/epoxy shell. The curvature at a specific compressive strain is obtained when the tensile and compressive forces in the section reach equilibrium. The stress state, and consequently the moment, is computed from the curvature and the constitutive relations. It should be noted that the carbon/epoxy shell has internal circumferential ribs to improve the force transfer between the concrete and the shell. To take into account the effect of the ribs, the hoop modulus of the carbon/ epoxy shell section was modified by determining an equivalent hoop modulus as a function of the modulus of the

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Table 3 Concrete compressive cylinder strengths in MPa (ksi) Pour number ( ) and location

Design strength

7 Day

28 Day

Day of test (DOT)b

Age on DOT in day

(1) (1) (1) (1) (2) (2) (3) (3) (4) (4)

34.5 34.5 34.5 34.5 34.5 34.5 34.5 34.5 41.4 41.4

31.4 33.3 31.4 33.7 32.8 34.4 37.4 34.7 37.8 39.1

45.1 41.0 40.4 40.7 41.4 44.1 42.9 39.2 39.7 42.5

50.2 43.9 43.9 43.4 43.3 46.3 43.9 40.0 40.8 44.5

47 47 47 47 38 38 36 36 21 21

a b

Footing #1 Footing #2 Footing #3 Footing #4 Pylon #1 Pylon #2 Pylon #1 Pylon #2 Load stub #1 Load stub #2

(5) (5) (5) (5) (5) (5) (5) (5) (6) (6)

(4.56) (4.85) (4.56) (4.89) (4.76) (4.99) (5.42) (5.03) (5.48) (5.67)

(5.81) (5.95) (5.86) (5.9) (6.0) (6.39) (6.22) (5.69) (5.76)a (6.17)a

(7.28) (6.37) (6.37) (6.29) (6.28) (6.71) (6.37) (5.8) (5.92) (6.45)

Tested at 14-day strength since test occurred at 21-day strength. Predicted per ASTM C918.

section without ribs, the hoop modulus of the ribs and the ratio between the areas with the unidirectional ribs and the sections without ribs. The hoop modulus of the unidirectional ribs was calculated, using the rule of mixtures, as, ET ¼ Ef V f þ Em V m ¼ 235ð0:65Þ þ 3:5ð0:35Þ ¼ 154 GPa ½22:3 msi

ð2Þ

where Ef is the modulus of the fibers, Vf is the fiber volume fraction, which is assumed to be 65% for the ribs, Em is the modulus of the resin or matrix, and Vm is the volume fraction of the matrix. Using the results from Eq. (2) for the ribs and the analytical value for hoop modulus from Table 2 (E1 = 90 GPa [13.03 msi]), the equivalent transverse modulus can be calculated as   A2 10:6  154 ET ¼ E1 þ  E2 ¼ 90 þ 117 A1 ¼ 104 GPa ½15:1 msi

ð3Þ

where, A1 is the nominal area of the pylon composite shell assuming a 9 mm (0.35 in.) thickness and a width of 13 mm (0.5 in.) representing the tributary area between ribs, and A2 is the effective area of the rib, 10.6 mm2 (0.01643 in.2). Results of the preliminary analysis using moment–curvature with increasing values of the axial load between 17,195 kN (3864 kips) and 20,470 kN (4600 kips) are shown in Fig. 10 using the analytical material properties for the FRP shell as listed in Table 2, and as modified above for the transverse direction, and an average concrete strength of 41.4 MPa (6 ksi). The moment–curvature analysis was conducted for the two critical sections up the pylon height (the connection/splice region and the typical section with the nominal reinforcing cage). The analytical results demonstrate that an increase in axial load causes a slight increase in the moment capacity while reducing the curvature and consequently the displacement capacity for both the splice and typical sections. A comparison of results obtained using the unmodified FRP properties in the transverse direction and those obtained using the approximation for the inclusion of ribs through Eqs. (2) and (3) indicates that the effects of the ribs provides only

a slight increase in the moment capacity of the sections but has more of an impact on the curvature and consequently the displacement capacity, especially for the splice section. The moment–curvature results also demonstrate that an increase in axial load results in an increase in the flexural capacity of the section. This effect is clearly shown for moment capacity as a function of axial load in Fig. 11 as a function of increasing curvature and in Fig. 12 at ultimate curvature. For the splice section, the three initial values of curvature, / = 0.01 1/m (0.00025 1/in.), / = 0.03 1/m (0.00075 1/in.), and / = 0.05 1/m (0.00125 1/in.), represent points in the linear elastic range of the specimen, and in the post-yield regime, respectively. For the typical section, only two values of curvature / = 0.006 1/m (0.00015 1/in.), and / = 0.008 1/m (0.0002 1/in.) are shown since the carbon/epoxy shell section behaves in a relatively linear elastic fashion with little curvature capacity. It can be seen that as axial load increases the moment at a specific curvature increases in both the splice and typical sections. Furthermore, Figs. 11 and 12 show that the average slopes of the curves are steeper for the splice sections, indicating that axial load has a higher effect on the splice section than on the typical section. It is also noted that the effects of the circumferential ribs is more noticeable at higher levels of curvature. It is of interest to consider the effect of FRP shell hoop modulus on the moment capacity and results of a study of the splice section using values of hoop modulus for the carbon/epoxy shell ranging from 48.3 GPa (7.0 msi) to 103.8 GPa (15.0 msi) are shown in Fig. 13 at a fixed value of curvature of / = 0.02 1/m (0.0005 1/in.). It is noted that an increase in hoop modulus, which corresponds to an increase in confinement of the concrete, results in a higher moment capacity, and that as hoop modulus (Eh) increases, the slope of the moment vs. axial load plot increases slightly. The results indicate that axial load has more of an effect at higher hoop moduli values. However, it can also be seen from Fig. 13 that as a section becomes sufficiently confined, the influence of the hoop modulus on the moment values is less significant.

Y. Van Den Einde et al. / Composites: Part B 38 (2007) 685–702

695

Curvature (1/in) 4

0

0.0004

0.0008

0.0012

0.0016

2.5x10

5

2x10 4

5

1.5x10 4

1.5x10

5

1x10

4

1x10

Moment (kips-in)

Moment (KN-m)

2x10

4

5x10

5000

Ribs 0 0

0.015

0.03

0.045

0 0.075

0.06

Curvature (1/m) Splice Section

Typical Section

P=17195 KN (3864kips) P=17800 KN (4000 kips) P=18690 KN (4200 kips) P=19580 KN (4400 kips) P=20470 KN (4600 kips)

P=17195 KN (3864kips) P=17800 KN (4000 kips) P=18690 KN (4200 kips) P=19580 KN (4400 kips) P=20470 KN (4600 kips)

Fig. 10. Moment–curvature response for varying axial load levels.

Axial Load (kips) 2000 4 2x10

2500

3000

3500

4000

4500

5000 5

1.7x 10

5

4

Moment (KN-m)

5

1.5x 10 4

1.6x10

5

1.4x 10

5

1.3x 10 4

1.4x10

5

1.2x 10

5

1.1x 10

4

1.2x10

4

1x10

10000

12000

14000

16000

18000

20000

1x10

5

9x10 22000

4

Axial Load (KN) Splice Section

Typical Section

φ=0.01 Ribs

φ=0.006 Ribs

φ=0.03 Ribs

φ=0.008 Ribs

φ=0.05 Ribs φ=0.01 No Ribs

φ=0.006 No ribs φ=0.008 No ribs

φ=0.03 No Ribs φ=0.05 No Ribs Fig. 11. Influence of axial load on moment capacity at specified curvature levels.

Moment (kips-in)

1.6x 10

1.8x10

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Axial Load (kips) 2500

3000

3500

4000

4500

5000

Moment (KN-m)

2x10

4

4

1.8x10

1.9x10

5

1.8x10

5

1.7x10

5

1.6x10

5

1.5x10

5

1.4x10

5

1.3x10

5

1.2x10

5

1.1x10

5

4

1.6x10

Moment (kips-in)

2000

4

2.2x10

4

1.4x10

4

1.2x10

10000

12000

14000

16000

18000

20000

22000

Axial Load (KN) Splice Section

Typical Section

φ at Mu Ribs

φ at Mu Ribs

φ at Mu No Ribs

φ at Mu No Ribs

Fig. 12. Influence of axial load on moment capacity at ultimate curvature.

Axial Load (kips) 4000

4200

4400

4600

4800

16000

5

1.4x10

φ = 0.02 1/m (0.0005 1/in)

5

5

1.3x10

5

1.25x10

14000

5

Eh=48.3 GPa (7.0 msi) Eh=62.1 GPa (9.0 msi) Eh=75.9 GPa (11.0 msi) Eh=89.9 GPa (13.0 msi) Eh=103.8 GPa (15.0 msi)

13000

12000 17000

18000

19000

20000

21000

1.2x10

5

1.15x10

Moment (kips-in)

Moment (KN-m)

1.35x10 15000

5

1.1x10

22000

Axial Load (KN) Fig. 13. Influence of FRP shell transverse (hoop) modulus a curvature of 0.02 1/m.

4.2. Prediction of force–displacement response In order to obtain a prediction of the force–displacement response for the pylon section it is essential that the magnitude of applied axial load is determined. Since the axial load in the Pylon Connection Test setup is applied passively, the actual value of axial load is found using strain measured on the strands prior to stressing. The peak strains can also be determined during stressing of the tendons, and just prior to testing, and from these levels the strain values in the strands after all losses can be estimated. The total percentage of losses was determined as the percent change in strain multiplied by

80%, which was the percentage of stress on the tendons. Using this procedure for each of the strands the average percent total loss in the tendons was calculated as 9.4%. This average percent loss was multiplied by the ultimate axial force capacity of a single strand (261 kN, 58.6 kips) to determine the axial force in one strand (184 kN, 41.4 kips). Finally, the total axial load applied by the four 27 strand tendons just prior to the test was 108 times the axial load per strand, which resulted in an axial load of 19,892 kN (4470 kips), which is closer to the upper limit of the estimates listed earlier. To determine the force–displacement prediction, it is necessary to make an assumption of the plastic hinge length and the development

Y. Van Den Einde et al. / Composites: Part B 38 (2007) 685–702

length in the splice region. Several recommendations listed by Priestley et al. [4] for plastic hinge length were evaluated for both typical reinforced concrete columns and columns retrofitted by composite shells with a gap region, and the average value from these assessments was used (0.9 m, 3 ft). The development length for the splice longitudinal bars calculated from AASHTO requirements [3] was greater than 0.9 m (3 ft). Since the splice bars extended 1.8 m (6 ft) above the pylon base, it was assumed that any section evaluated between the footing and 0.9 m (3 ft) above the footing followed the splice connection moment–curvature response and any section above 0.9 m (3 ft) followed the typical section moment– curvature response. Although bars were present in the pylon from 0.9 to 1.8 m (3–6 ft), for simplification in the analysis they were assumed to be undeveloped and would hence not contribute any strength. The pylon was discretized up the height using 0.15 m (6 in.) intervals in the plastic hinge region from 0 to 0.9 m (0–3 ft), and 0.3 m (1 ft) intervals throughout the rest of the height of the pylon. Using the axial load value of 19,892 kN (4470 kips), the analytical properties for the shell with modifications to include the effect of the ribs in the hoop modulus, moment–curvature analyses were conducted for both the splice and typical sections. To determine the structural force–displacement profile from these sectional moment– curvature responses for a given lateral load, the moment demand at each section can be calculated assuming that moment varies linearly down the pylon. Using these moment values and the appropriate moment–curvature curve, which depended on the location up the pylon height of the section being evaluated, the corresponding curvatures can then be determined. The curvatures can in turn be integrated up the entire height of the pylon to get rotations. Similarly, the rotations can be integrated up the height to get displacements. This procedure can be per-

697

formed for increasing values of lateral loads to obtain the full force–displacement profile as shown in Fig. 14. For ease of reference the Extreme (seismic) (1222 kN, 275 kips), maximum Service (395 kN, 89 kips), and maximum Strength (622 kN, 140 kips) demand levels, are also shown in the figure. This curve can be used approximated through use of bilinearity to determine the ideal yield force, Fy, for testing by at a concrete compression strain of 0.004. It is pointed out that the initial non-linearity in Fig. 14 in what is typically the linear elastic regime is due to the extreme axial load on the pylon which causes the concrete strains to appear to be non-linear even though the steel strains are relatively low. From this force–displacement analysis, under the seismic demand level, the longitudinal strain levels in the composite are estimated to be ec = 0.00248 in compression and et = 0.00289 in tension. These are below the pre-set design strain limits of ec = 0.0045 and et = 0.0075, respectively, assuming / = 0.75, and are also well below the compression and tension strain limits of ec = 0.003 and et = 0.005, respectively, if / = 0.5 is assumed for conservatism. 5. Test results and discussion The experimentally obtained hysteresis curves for the test are shown in Fig. 15. The test exceeded all requirements for displacement capacity and failed during cycling at displacement ductility 7, which was six times the displacement required under the maximum design demand for the bridge. A comparison of the predicted force–displacement response envelope using the moment–curvature analysis (with termination of the predicted response based on the shell reaching the ultimate compression strain limit) described earlier is also shown for comparison. Although damage such as crushing of the cover and core concrete, and buckling of the longitudinal splice

25000

Experimental Assuming Pure Cantilever Experimental with Correction for Overturning Effects Analytical

Base Moment (kN-m)

20000

15000 Seismic Demand: 12566 kN-m

10000

Strength Demand: 6401 kN-m

5000

Service Demand: 4061 kN-m

0 0

100

200

300

400

Displacement (mm) Fig. 14. Moment–displacement profile.

500

600

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Displacement (in) -400

-200

0

200

400

Actuator Load (kN)

2000

400

Extreme I

1000

200

Strength I Service I

0

0

-200

-1000

Analytical Experimental

-2000 -600

600

-400

-200

0

200

400

Actuator Load (kips)

-600

-400

600

Displacement (mm) Fig. 15. Experimental force–displacement hystereses plot.

reinforcement in the connection region occurred during the test, up until displacement ductility 3, damage to the footing due to strain penetration of the longitudinal splice reinforcement was minor and could be considered cosmetic. The ultimate capacity of the specimen occurred at displacement ductility 7, which corresponds to a 5% drift ratio or 512 mm (20.2 in.) of lateral displacement at the top, as depicted in Fig. 16. It was noted that during the first pull cycle to displacement ductility 7, five of the longitudinal bars on the push side of the column ruptured. During the second push cycle to displacement ductility 7, eight of the extreme longitudinal bars that had previously buckled on the pull side of the pylon finally ruptured. During the final pull cycle, several more of the buckled bars on the push side of the pylon ruptured bringing the total on this side to seven bars. At the end of the test, 15 of the 30 longitudinal bars in the splice region had fractured, but the overall global capacity of the specimen only degraded approximately 20%. The concrete core at the end of the test was crushed significantly. Fig. 17 shows the ruptured longitudinal bars on both the push and pull sides of the Pylon after the test.

5.1. Consideration of variation in applied axial load during testing Although the axial load at the beginning of the test was 19,892 kN (4470 kips), as the pylon was loaded cyclically, the value of the axial load changed. The gain in strength of each tendon with increasing lateral load was compensated with a loss or rupture of some individual wires at the last levels of ductility. Using the experimentally measured peak strains in the tendons the percent change in strain between the peak tension and compression strains measured during the test and the strain in the tendons

Fig. 16. Photograph of pylon at extreme drift during the push cycle at displacement ductility 7.

found prior to testing can be used to determine the average percent change in both tension and compression strain for all gaged strands (%Detension, %Decompression) These can then

Y. Van Den Einde et al. / Composites: Part B 38 (2007) 685–702

Fig. 17. Ruptured reinforcing bars at the end of the test: (a) push side of pylon; (b) pull side of pylon.

be used to determine the increment in axial load from the start of the test to displacement ductility 7 as DP ¼

1 1  P total;ini  %Detension þ  P total;ini  %Decompression 2 2

ð4Þ

which results in a load of 2394 kN (538 kips). During the test, a number of popping sounds were heard indicating rupturing of individual wires. Given that ten total strands (70 wires) were lost during the test, the total axial load at the end of the test can be found as 98 times the initial force per strand (184 kN, 41.4 kips) plus DP from Eq. (4), resulting in a total axial load of 20,261 kN (4553 kips). It can be shown that this slight variation in axial load had a relatively minor effect on overall capacity. 5.2. Overturning moment effect Another issue with the passive tendon system is that the configuration of the tendons causes a restoring force on the test specimen, which counteracts the force applied by the actuator, resulting in measured force levels in the actuator that were not indicative of the true applied force on the specimen. Rather than a triangular moment distribution down the pylon expected of a cantilever, the overturning moment induced by the tendons thus creates a negative moment at the top of the specimen and a considerably lower moment at the base of the test specimen. The overturning moment can be calculated using strain measured by gages placed on the tendons. During loading of the test specimen, all of the tendons experienced an increase in strain. However, the two tendons on the tension side of the pylon saw a significantly larger increase in strain than

699

the other two tendons on the compression side of the pylon. At various load levels, the average strains were determined for both the elongating tendons and the tendons experiencing less elongation. The change in force due to this differential strain between the two sides of the tendons can be calculated and from this force couple, the corresponding overturning moment at each load level can be determined and used to modify the cantilever moment due to the applied actuator force to get the true base moment. The force–displacement prediction described earlier can then be modified to take into consideration the overturning effects in order to directly compare experimental and analytical results. Essentially, for various applied actuator force levels, the corresponding moments at incremental heights up the pylon need to be determined not by including the overturning moment effects. These overturning effects were found to be bilinear throughout the response of the test. For low levels of loading, the amount of overturning moment increased with a shallow slope. However, at applied actuator force levels above 1500 kN (337 kips) the slope of the line increased considerably such that with a small increase in actuator force, the amount of overturning moment increased substantially. It should be noted that including the overturning effects did not change the analytical prediction significantly. Since the pylon test was no longer a pure cantilever test, the results of this assessment are shown as moment at the base of the pylon vs. top displacement rather than force at the pylon top vs. top displacement (see Fig. 14). From Fig. 14 it is evident that including the overturning moment in the experimental results has a significant effect on the overall response dropping it to levels close to the analytical prediction, which is based on moment–curvature analyses of the critical pylon sections. When overturning moment effects are taken into consideration, the actuator force levels that correspond to the Service, Strength, and Seismic moment demand levels increase from 395 kN (89 kips) to 408 kN (92 kips), 622 kN (140 kips) to 656 kN (147 kips), and 1222 kN (275 kips) to 1303 kN (293 kips), respectively. This is important in the evaluation of strains at the various demand levels are evaluated. It is emphasized that the design objective of ensuring that the initiation of plastic hinge formation at the splice connection region occurred at a curvature ductility level of 1.5 or less was achieved despite the fact that the seismic demand level occurred at 1303 kN rather than 1222 kN as assumed without overturning effects. The actual displacement ductility level for the seismic demand level, which is a measure of overall global structural response, was 1.3 and the corresponding curvature ductility, which is a function of section response, was slightly less. The pylon connection design objective is based on curvature ductility since it is easier to evaluate during design using moment– curvature analysis tools. The initiation of the plastic hinge slightly earlier than designed for is conservative since plastic behavior occurs substantially before any critical strain levels in the composite jacket are reached. Also, the

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moments in the pylon do not impact other composite elements since the pylon transfers all of its loads to the conventionally designed connection. 5.3. Longitudinal strain results Although several strain gages placed on the steel reinforcement were lost prior to testing, and a number were lost after reaching a displacement ductility level of 3, sufficient gages remained to indicate that the maximum recorded strains prior to the loss of gages in compression and tension was 0.0043 and 0.0029, respectively. The strains in the splice steel were at or below yield under the seismic load case. A comparison of the FRP composite compression and tension experimental strains, the strains obtained from the moment–curvature analysis taking overturning effects and the variation in axial load during the test into account, and the strain limits set forth as design criteria are listed in Table 4. It is noted that the correlation between the analytical and experimental strain values for the carbon/epoxy composite shell is poor especially for

the tension strains, underlining the need for further research in order to improve the analytical model used for moment–curvature analysis. However, the measured strain levels in the carbon/epoxy composite shell at the Extreme demand level, which were ec = 0.0019 and et = 0.0011 in compression and tension, respectively, are noted to be well below the 50% allowables of ec = 0.003 and et = 0.005. Furthermore, the maximum experimental strains observed in the composite shell occurred at a height of approximately 1829 mm (6 ft), which was the location where the longitudinal steel bars in the splice region were terminated. The low experimental strains in the carbon/ epoxy composite shell at the Service, Strength, and Seismic demand levels indicate that the shell is not being activated at these load levels. To evaluate the transfer of strains up the height of the pylon, the longitudinal strains in the FRP shell (the jacket) are plotted against the strain results from the corresponding gages on the longitudinal splice steel bars for both the push and pull loading directions (Figs. 18 and 19, respectively). The results show that for

Table 4 Comparison of longitudinal FRP composite shell strains Level

Analytical strainsa

Experimental strains

Allowable strains

ec

et

ec

et

ec

et

Service

0.00065

0.00002

0.00069

0.00002

Strength

0.0009

0.00026

0.00091

0.00011

Seismic

0.00248

0.00289

0.0019

0.0013

Ultimate

0.00372

0.0051

0.0043

0.0029

=(0.3)(ecu)b 0.0018 =(0.5)(ecu)c 0.003 =(0.5)(ecu)d 0.003 0.006

=(0.3)(etu)b 0.003 = (0.5)(etu)c 0.005 =(0.5)(etu)d 0.005 0.01

b c d

From moment–curvature analysis. / = 0.3 in Design criteria. / = 0.5 in Design criteria. / = 0.75 in Design criteria but assumed as / = 0.5 in this document for conservatism.

2500 Level 5 Level 6 Level 7 Level 8 Reference

2000

Jacket (microstrain)

a

1500

Push N

S

1000

500

Level 8 Level 7 Level 6 Level 5

0

-500 -500

0

500

1000

1500

2000

2500

3000

3500

Steel (microstrain) Fig. 18. Comparison of strains in steel reinforcement and FRP shell in the push direction.

Y. Van Den Einde et al. / Composites: Part B 38 (2007) 685–702

701

2500

Pull

Jacket (microstrain)

2000

Level 5 Level 7 Reference

1500

S

N

1000

500 Level 7 Level 5

0

-500 -500

0

500

1000

1500

2000

2500

3000

3500

Steel (microstrain) Fig. 19. Comparison of strains in steel reinforcement and FRP shell in the pull direction.

Level 5, which is located 483 mm (19 in.) above the top of the footing, the steel strain is greater than the FRP shell strain indicating that the longitudinal steel reinforcement is providing more of the flexural capacity of the section. However, at Levels 6 and 7 (see Fig. 18) the ratio between the strains is close to unity demonstrating that the longitudinal steel has successfully transferred its strains and forces to the FRP shell. This successful force transfer occurred within 724 mm (28.5 in.) and 965 mm (38 in.) above the footing. At Level 8 (1448 mm, 57 in. above footing) the jacket strains are considerably higher than the steel strains. At this location, the FRP shell has already been fully developed and the steel reinforcement, which terminates at approximately 1829 mm (72 in.) above the footing, has not developed to its fully capacity. These results indicate that since only 724 mm (28.5 in.) is required to develop the pylon shell to its fully capacity, the use of 1829 mm (72 in.) for splice length of the steel reinforcing bars is extremely conservative, and that essentially, beyond the critical lag length for the FRP shell, the steel is no longer being utilized. However, Figs. 18 and 19 also indicate that it took the splice steel approximately 864 mm (34 in.) to develop fully from the top of the splice reinforcement down towards the footing. This was determined by subtracting the termination location of the splice steel reinforcement at 1829 mm (72 in.) above the footing by the height at Level 7 (965 mm, 38 in.) where the steel successfully transferred its forces to the shell. This development length corresponds closely to the development length of 850 mm (33.5 in.) as recommended by AASHTO, which is a function of the area and yield strength of the longitudinal bar and the concrete compression strength [3]. It could be argued that the total lap splice length should extend one full development length beyond the lag length of the shell. This would require a total splice length for the longitudinal bars measured from the footing of 1588 mm (62.5 in.), which is approximately

equal to the diameter of the pylon. Although this appears conservative since the carbon/epoxy shell and the longitudinal splice steel reinforcement both contribute to the longitudinal stiffness of the pylon, even if only partially developed, it is considerably less than the 1829 mm (72 in.) provided in the test. 5.4. Transverse strain results Transverse strains in both the FRP composite shell and steel hoops were evaluated to better understand the confining effects of the composite shell, especially in the gap region. It was seen that for the most part, the transverse strains in the FRP shell and steel were similar to each other up to the Extreme level of loading at which point the transverse strains in the carbon/epoxy composite shell were approximately 25% higher. This indicates that once the steel yields, the confinement of the concrete in the pylon is dominated by the carbon/epoxy composite shell. It is noted that the side of the pylon that was subjected to compression, which varied depending on whether the pylon was pushed or pulled, registers higher transverse strains. Table 5 shows the values of the FRP and steel strains at the first cycle of the various load levels in comparison with the limits set forth as design allowables. The strain values are Table 5 Comparison of transverse strain comparison in the FRP composite shell and in reinforcing steel Level

Experimental jacket strain

Experimental steel strain

Allowable strain

Service Strength Seismic lD = 5 Ultimate

0.0003 0.00032 0.00054 0.0041 0.0048

0.00019 0.00026 0.00045 0.003 –a

0.0018 0.003 0.003 – 0.006

a

Steel strain gages were lost after lD = 5.

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based on the corrected loads for the Service, Strength, and Seismic demand levels after taking the overturning moment contribution into effect. The maximum tensile transverse strains in the FRP were approximately 0.0041 up to a displacement ductility level of 5. In the steel, the transverse strains were well below yield at this extreme demand level, and were less than 0.003 at displacement ductility 5. On the tension side of the pylon, the transverse strain levels reduced significantly in both the composite jacket and the steel. Throughout the entire loading history, the transverse strain values in the composite shell were well below the all of the design limits. This demonstrates that the amount of confinement provided by both the steel hoops and the FRP shell was sufficient since the transverse strains at high ductility levels were still very low. 6. Summary and conclusions The results indicate that the Pylon Connection Test met all requirements for a successful proof test. The force capacity of the test unit was significantly greater than the maximum expected force demand and the displacement capacity was close to six times the displacement observed under the Extreme demand level. Taking into consideration the overturning moment effects, the Extreme demand level corresponded to displacement ductility of approximately 1.3, which demonstrates that the objective of designing an essentially elastic splice connection with ductility of 1.5 or less under the Extreme load combination can be successfully achieved using a FRP shell based Pylon system. The longitudinal compression and tension strain levels in the composite shell were seen to be well below the very conservative design allowables. Furthermore, a comparison between strains in the longitudinal steel and FRP shell showed that the development length provided for the steel in the splice region was more than sufficient in transferring the forces to the composite shell. The force transfer length between the steel and the jacket was approximately 724 mm (28.5 in.). The development length of the steel bars was found to be approximately 864 mm (34 in.), which is very close to the value assumed in the design based on AASHTO recommendations for a 29 mm (#9) diameter bar. The results thus indicate that the splice length provided in the connection region could be reduced substantially. It is suggested that lap splice lengths provided in the connection regions for the pylon should be greater than or equal to one pylon diameter (1524 mm, 60 in.), which corresponds to the sum of the development lengths of the shell (724 mm, 28.5 in.) and splice steel longitudinal bars (864 mm, 34 in.). It is also recommended that the steel splice reinforcing bars be terminated in a staggered fashion

so that the change in capacity between the splice connection region and the typical pylon section does not occur abruptly. As expected, the amount of transverse reinforcement provided by the FRP composite shell and steel hoops in the splice region was found to be conservative since transverse strains up to a displacement ductility level of 5 were well below the allowables. The confinement from the shell and hoops was able to resist pullout failure of the lap splice longitudinal bars and sufficiently confine the plastic hinge region for adequate ductility capacity. Although continued experimentation and analyses are required to better understand the effects of confinement, the conservatism indicates that perhaps the amount of hoop fibers in the shell throughout the height of the composite component could be reduced in order to optimize the design and lower costs. Acknowledgements The authors acknowledge the California Department of Transportation (Caltrans), the Federal Highway Administration (FHWA), the State of California, and the University of California, San Diego (UCSD) for funding the research conducted in this report. The extensive time and assistance that the Dywidag Systems International (DSI) technicians provided during the placement and stressing of the axial load prestressing tendons is greatly appreciated. References [1] Starossek U. Cable-stayed bridge concepts for longer spans. ASCE J Bridge Eng 1996;1(3):99–103. [2] AASHTO, AASHTO LRFD Bridge Design Specifications Modifications to Chapter 11 (abutments, walls, earth pressure), 2nd ed. Washington (DC): American Association of State Highway and Transportation Officials; 2001. [3] AASHTO (including 1999 and 2000 interims), ‘‘AASHTO LRFD Bridge Design Specifications, 2nd ed. Washington (DC): American Association of State Highway and Transportation Officials; 1998. [4] Priestley MJN, Seible F, Chai YH. Design guidelines for assessment retrofit and repair of bridges for seismic performance. Structural Systems Research Project, SSRP-92/01. San Diego, La Jolla: University of California; 1992. August. [5] Burguen˜o R. Ductile and system characterization of modular fiber reinforced polymer (FRP) short- and medium-span bridges. Department of Structural Engineering, University of California, San Diego, Ph.D. Dissertation, La Jolla; 1999. [6] Karbhari VM, Seible F, Burgueno R, Davol A, Wernli M, Zhao L. Structural characterization of fiber-reinforced composite short- and medium-span bridge systems. Appl Compos Mater 2000;7(2/3): 151–82. [7] Mander JB, Priestley MJN, Park R. Theoretical stress–strain model for confined concrete. ASCE J Struct Eng 1988;114(8):1804–23.