Design and construction of caisson breakwaters — the Japanese experience

COASTAL ENGINEERING Coastal Engineering,22 (1994) 57-77

Design and construction of caisson breakwatersthe Japanese experience Katsutoshi T a n i m o t o a and Shigeo Takahashi b ~Saitama Universit3'. 255 Shimo-okubo. Urawa, 338 Japan bPort and Harbour Research Institute, I-1-3, Nagase, Yokosuka. 239 Japan

( Received 15 September 1992; accepted after revision 15 June 1993)

Abstract The history of breakwater construction in Japan is reviewed and present design methods of vertical breakwaters are discussed in terms of wave forces and breakwater stability. Current design techniques of vertical breakwaters in Japan are the culmination of many years of experiences and practical research in this field. The Goda formula, which accurately predicts the design wave forces under most conditions of vertical breakwaters, is an excellent aid in the design of the breakwaters. Under certain conditions, excessive impulsive forces due to the action of breaking waves are exerted on the upright section. Such conditions must be minimized in the design of vertical breakwaters, and the adoption of vertical breakwaters covered with wave dissipating concrete blocks or the wave dissipating caisson breakwaters offer potential solutions of this problem. The bearing capacity of rubble mound foundation and the stability of armor units are also discussed.

1. Introduction Since Japan is surrounded by rough seas, breakwaters are of great importance in the development and expansion of harbour facilities, and so great emphasis is placed on engineering in terms of both design and construction costs. As a result, intensive studies of wave actions on breakwaters and structural performances have been carried out by various organizations related to harbour engineering. The most common structural type of breakwaters in Japan is a composite (mixed) type, which consists of a rubble mound foundation and an upright section, while rubble mound breakwaters are currently the most common type worldwide. A typical cross-section of the composite type breakwater is shown in Fig. 1. It may be called a vertical breakwater, although it is applied not only to standing wave conditions but also to breaking wave 0378-3839/94/$07.00 © 1994ElsevierScienceB.V. All rights reserved S S D I 0 3 7 8 - 3 8 3 9 ( 9 3 ) E0044-J

58

K. Tanimoto, S. Takahashi / Coastal Engineering 22 ( I 994) 57-77

S.W.L

armour

v

/

J , ; o.

,-~

,:~t~

~7

stones ~ _ L _ r ~ _ Z ~ \ ~ L m _ ~

ss°n

\ rubble

mound

foundation

Fig. 1. Standard cross section of caisson breakwater. Rumoi

SEA OF JAPAN

i

/~IF"

\

Funakawa ~C~

x'~__c~"°

=

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~

/

o /

s~k,,,_...j .o Sakai

~

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-

~-'Karnaishi

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,, . neaono

Om

L"

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\ Miyazaki

Nagashima

//

1¢°°F

pAc,F,cOCEAN

Fig. 2. Map of ports referred to in the paper. conditions. Composite type breakwaters whose upright sections are covered with wave dissipating concrete blocks are also common in Japan. In the present paper, the history of breakwater design and construction in Japan is reviewed (Tanimoto and Goda, 1992) and present design methods of vertical breakwaters are discussed in terms of wave forces and overall stability. Fig. 2 indicates the location of ports which are referred to in this paper.

2. Historical development of breakwaters in Japan 2.1. Historical breakwaters

The first modern breakwater in Japan was constructed at Yokohama Port in 1891, and was designed by H.S. Palmer, a retired British Major-General. A cross-section is shown in Fig. 3. Palmer introduced the technology of relatively low mound vertical breakwaters as shown in the figure. In those years, the rubble mound foundation of composite type breakwater became gradually lower and lower in Britain. The Yokohama breakwater, having a total length of 3630 m, was first completed in 1896. It was redesigned, however, with concrete blocks for the whole upright section after it was damaged by a storm in 1902.

K. Tanimoto, S. T~tkahashi I Coastal Engineering 22 (1994) 57-77 6.71

,

.2_H.W.L + 2 . 0 4 + 2 . 3 $ - [ ~ _~_L.W.L + 0.00 ~ _:~2.84 ~

S

g

59

.~ /,Concrete in Bags

_no_Of i_t Clayey S t o n e s ~ _

l~Grgun_d ........

Rubble(0..5~2.0ton) Fig. 3. Original cross-section of east breakwater before damage at Yokohama Port (1891-1896).

67 Zl-~/P°ured ~ X363 Concrete-fill +1 wLW.OS.T +0.61 ' ~ ( ' ÷

±o.oo 7

a ,5 II-":t~ll~ -3.03 -5.45 ~-,,.. . ~ -

-5,5

Engineering Fill~

Fig. 4. Cross-section of island breakwater at Otaru Port (1912-1917).

~.8. I Concrete *3.b

~..H.W. L + 2.0 ~

V L.WL-+O.O

.

+^~ ~ . L ._

/¢'2Y

--I ~ - - - ' ~ -5.2

_

~ ~ R u b b l e

~

~

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~

~

Clay Fig. 5. Cross section of outer breakwater at Yokohama Port (1928-1943).

Construction of the first vertical breakwater consisting of concrete blocks, built at Otaru Port by I. Hiroi, lasted from 1897 to 1907. Hiroi used the breakwaters at Karachi, Madras, and Colombo as prototypes of the Otaru breakwater. The breakwater with reinforced concrete caissons in Japan was introduced at Kobe Port for the first time in 1911, and then utilized for an island breakwater at Otaru Port in 1912. Since the wave conditions are much more severe at Otaru than Kobe, the breakwater was quite sturdy, as shown in Fig. 4. The caissons weighing 883 tons were filled with poured concrete. This breakwater still exists today as it was constructed. The present design wave conditions are 4.8 meters for the significant wave height, 8.2 m for the maximum wave height, and 11.5 s for the significant wave period for very oblique wave incidence. While concrete was the principal fill material for the early caisson cells, gravel and sand have gradually replaced concrete. One of the first examples of a sand-filled breakwater is the outer breakwater at Yokohama Port, which was designed by S. Samejima; its construction lasted from 1928 to 1943. As shown in Fig. 5, the caissons were very sturdy, with a

60

K. Tanimoto. S. Takahashi I Coastal Engineering 22 (1994) 57-77

mixture of sand and cobbles to give the greatest possible density. Filling of caissons with sand was initially done in areas where the wave conditions were relatively mild, but sand fill soon became popular in areas of rough seas as well. This was particularly so in the period shortly after World War II, when cement was scarce because of damage to production facilities. 2.2. Caisson breakwaters' Caisson breakwaters soon became the predominant type for use in rough seas. It is said that the early introduction of reinforced concrete caissons influenced the later development of vertical breakwaters in Japan. Recently, the Construction Office of Port Onahama, a major port in northern Japan, celebrated their 1500th caisson fabricated in the construction yards of the port. The first caisson was built in 1932, 131 caissons were made in a year of 1971 and more than 20 caissons are being constructed every year still now. Those caissons were used for breakwaters and bulkheads of not only Onahama Port but also nearby ports. The production level at Onahama Port is typical of the rapid development of Japanese ports in general, and especially of caisson breakwaters. The widest caisson breakwater is found at Hedono Port, located on a southwest island. The caisson is 38 m wide (Fig. 6) based on the following design wave characteristics: 9.7 m significant wave height, 17.5 m maximum wave height, and 13.2 s significant wave period. In 1992, the longest caisson was used as a temporary breakwater at Kochi Port: One unit of the caissons is 100 m long, 19.7 m wide, and 13.5 m high. The caisson was designed to incorporate steel frames and prestressed concrete walls. The caisson was fabricated in a ship dock and towed to the site over a distance of 370 km. Fig. 7 shows the long caisson installed at Kochi Port in 1992. The long caisson is expected to receive a reduced mean wave force per unit length for obliquely incident waves because of the phase difference along the extension. The deepest caisson breakwater in Japan, and probably in the world, is under construction at a maximum depth of 60 m at Kamaishi Port. Fig. 8 shows a cross-section at the deepest part. The shape of the lower part of the upright section is trapezoidal to provide a wide stable bottom slab. This caisson is a wave dissipating type with double horizontal slit walls at the upper part. Since 1961, when the first perforated caisson was proposed by G.E. Jarlan, such wave-dissipating caisson breakwaters have been constructed at many ports in Japan. ,

+o.ooF-

38.0

--

- -

Fig. 6. Cross-section of the widest caisson breakwater at Hedono Port.

K. Tanimoto, S. Takahashi / Coasud Engineering 22 (1994) 57-77

61

Fig. 7. Long caisson in Kochi Port (courtesy of 3rd Port Construction Bureau. MOT). 16D,

~ 0

~7 I:I.W L + 1.50

.~n

l,..~r~.~

"~'~-

L w,±ooo

T

- -

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- z~'--'-~'~-~l

~,;.

~I~)-IN~-' ~u ~

-6o.o

-~"["~i - - '-,.,,..

......

Fig. 8. Cross-section of the deepest caisson breakwater at Kamaishi Port. 1 2 . ~ 0 Rubble Stone Fill V g -

H.W.L .w.

I. 15

0.4!,~"L_~

~

:ooo

- ~b.O

Fig. 9. Cross-section of tsunami protection breakwater at Ofunato Port ( 1963-1967 ).

2.3. Tsunami and storm surge protection breakwaters Breakwaters are also constructed as barriers against tsunamis and storm surges. Storm surge protection breakwaters, with a total length of 8,250 m, were built at Nagoya Port from 1962 to 1964 after a catastrophe caused by the Isewan Typhoon in 1959. Since sea bottom in the area is soft ground consisting of very low-strength clayey soil deposits, a sand drain method was used to stabilize the soil. Subsequently, a tsunami protection breakwater as shown in Fig. 9 was built at Ofunato

62

K. Tanimoto, S. Takahashi / Coastal Engineering 22 (1994) 5 ~ 7 7

Port from 1963 to 1967 after the damage sustained due to the Chilean Earthquake Tsunami in 1960. The breakwater is located at the bay mouth, where the maximum water depth is 38 meters. At the central harbour entrance with an opening of 200 m, a submerged dike up to the level - 16.3 m was built to reduce the opening area as much as possible. The Kamaishi breakwater is also a tsunami breakwater located near Ofunato Port.

2.4. Caisson breakwater covered with wave-dissipating concrete blocks The emergence of artificial concrete blocks called tetrapods, which was proposed by P. Danel in 1949, greatly affected the design of breakwater structures in Japan and this type of blocks was first used at Yagi Port in 1955. Thereafter, a caisson breakwater at Rumoi Port, which was originally built from 1911 to 1933 and occasionally suffered damage over the years, was reinforced with artificial concrete blocks (Fig. 10). This type of breakwater came into widespread use as a breakwater in relatively shallow areas with high breaking waves. The main function of artificial concrete blocks in front of a vertical wall is to reduce the impact of waves on the wall, wave overtopping, and wave reflection by dissipating wave energy, while the artificial blocks for rubble mound breakwaters are used to protect the sublayers. Consequently, they are not called armor blocks, but "wave-dissipating concrete blocks" in Japan.

2.5. New caisson breakwater design Various new caisson structures which are suited to varying water depths and wave conditions have been developed in Japan, i.e. a curved slit caisson, a multi-cellular caisson (Okada et al., 1990), a dual cylindrical caisson, a semi-circular caisson, and a wave power extracting caisson (Takahashi, 1988), etc. Some of these new designs have already been constructed at several ports and other prototypes are undergoing field tests. 15.O

j 8.0•I 106 . I Tetropods +4,5 +3'016 i LWL ~-00 ~ - 0 2 5

Fig. 10. Cross-section of caisson breakwater reinforced with wave dissipating concrete blocks at Rumoi Port.

18.0

~, LW.L_+0.0

A~;~H,W.L+u.~

{'4"'5/' • +4"~ ~-50 ~ '

Fig. I 1. Cross-section of curved-slit caisson breakwater at Funakawa Port.

K. Tanimoto, S. Takahashi / Coastal Engineering 22 (I 994) 57-77

Fig. 12. Conceptual figure of dual cylindrical caisson breakwater.

Fig. 13. Dual cylindrical caisson in Nagashima Port (courtesy of Mie Prefectural Government).

63

64

K. Tanimou~, S, Takahashi / Coastal Engineering 22 (1994) 57 77 ~- K . O

_

LWL*O005 -5.5~0H,W.L* . 5 ~ _ _ A s halt -5.5Mat v-85 . . . . . . . .

-7."

~ . 4.~.". . . "qi . -8.5 . . . . .. . ... . . ,,~Li~ . 75 £85 ~...-.-... -7~5 t zo ~,oI 90 I ,4o _jeo_13d_5ol . . . . . .

Fig. 14. Possible cross section of semi-circular caisson breakwater for extremely high breakers.

Fig. 15. Semi-circular caisson in Miyazaki Port (courtesy of 4th Port Construction Bureau, MOT).

The curved slit caisson was proposed to apply a wave-dissipating caisson for rough seas in 1976. After successful laboratory and field tests, the first curved slit caisson breakwater 150 m long was constructed at Funakawa Port in 1984. A cross-section of this caisson breakwater is shown in Fig. I 1. The curved-slit members were prefabricated with prestressed concrete and fixed to the main body of the caisson by a dry joint method. According to calculations (Tanimoto et al., 1987), trapezoidal and cylindrical caissons are more effective and stable as the water depth increases and in rough seas. The Kamaishi breakwater is a typical example of a deep water trapezoidal caisson. Even for deep water breakwaters, the construction of low reflective and permeable structures is sometimes preferable to minimize the environmental impact. In response to this consideration, a dual cylindrical caisson (Fig. 12) has been developed. A doughnutshaped wave chamber is formed between the outer permeable cylinder and the inner impermeable cylinder. Field tests have been conducted at Sakai Port from 1989 (Tanimoto et al., 1992), and a 180-m breakwater was constructed in a marine recreational area of Nagashima Port (Fig. 13), although the water depth is not so large ( 11 meters). For shallow water, a semi-circular caisson breakwater is under development at Miyazaki Port. Because of the high stability against waves and the soft feature with a round top, it is

K. Tanimoto, S. Takahashi/ Coastal Engineering 22 (1994) 57-77

65

expected that a semi-circular caisson is well suited for offshore breakwaters designed to protect beaches from erosion, particularly in marine recreational areas. Fig. 14 shows a cross-section of design under consideration for use in areas with very intensive breaking waves. Fig. 15 shows a semi-circular caisson manufactured in a caisson yard at Miyazaki Port.

3. Wave forces and stability examinations 3.1. Design standards for port and harbour facilities in Japan

The various design of breakwaters in Japan have been improved based on experiences in the field and extensive research. Those engineering experiences have continuously been integrated into the design manuals and standards for port and harbour facilities by the Port and Harbour Bureau, Ministry of Transport (Port and Harbour Bureau, Ministry of Transport, 1989). The first design manual for wharves and breakwaters was published in 1950 and revised in 1959. In 1967, the first edition of design standards for port and harbour structures was published and have been revised several times since then. The first edition of technical standards with expository comments for port and harbour facilities was published in 1980 and last revised in 1989. Goda's formula to calculate the design wave forces on the upright section was adopted as a standard formula in the design of vertical breakwaters in the 1980 edition, and a new method to examine the bearing capacity of rubble mound foundations was introduced in the 1989 edition. In this chapter, the Goda pressure formula, and other methods of stability analysis of vertical breakwaters against waves are outlined. Problems related to excessive impulsive forces by breaking waves are discussed in the next chapter. 3.2. Wave forces on the upright section

External wave forces on the upright section are the most important considerations in the design of vertical breakwaters. Therefore, these forces have been the subject of intensive research. As a result, a formula to calculate the design wave forces was established by Goda in 1973. With a later modification to account for the effect of oblique wave incidence, this formula has been successfully applied in the design of vertical breakwaters in Japan. In the Goda formula, the wave pressure along a vertical wall is assumed to have a trapezoidal distribution both above and below the still water level, and the uplift pressure acting on the bottom of the uptight section is assumed to have a triangular distribution (Fig. 16). The buoyancy is calculated for the displacement volume of the upright section in still water below the design water level. In the figure, h denotes the water depth in front of the breakwater, d the depth above the armor layer of the rubble mound foundation, h' the distance from the design water level to the bottom of the uptight section, and hc the elevation of the breakwater above the design water level. The elevation at which the wave pressure is exerted, 77* and the representative wave pressure intensities p~, P2, P3 and Pu can be written in a generalized form as follows:

K. Tanimoto, S. Takahashi/ Coastal Engineering 22 (1994) 57-77

66

r/* = 0 . 7 5 ( 1 + cos/3)A jHD

(1)

Pl = 0 . 5 ( 1 + COS/3)(Alal + A2 a2COS2/3)w~rHD

(2)

p3 = ce3pj

(3)

P4 = a4Pl

(4)

p. = 0 . 5 ( 1 +cos/3)A3ch o~3woH D

(5)

in which, oq = 0.6 + 0.5 [ ( 4 r r h / L D ) / s i n h ( 4 7 r h / L D ) ] 2

(6)

cr+_= min { ( 1 - d / h b ) ( H o / d ) 2 / 3 , 2 d / H D }

(7)

ce3 = 1 - ( h ' / h ) [ 1 - 1 / c o s h ( 2 r r h / L D ) ]

(8)

c~4 = 1 - h<*/r/*

(9) (lO)

h* = min{ r/*,h,. }

/3 is the angle between the direction of wave approach and a line normal to the breakwater, A ~, A 2, A 3 are the modification factors depending on the structural type, Ho, Lo are the wave height and the wave length applied to the calculation of design wave forces, respectively, wo is the specific weight of sea water, hb is the water depth at the location offshore by the distance of five times the significant wave height H~/3, min{ a, b} is the minimum of a and b. For the ordinary vertical breakwater, the modification factors of A j = A~_= h3 = 1 are applied since the Goda formula was proposed originally to describe this type of breakwater. The first factor A ~ represents the reduction or increase of the slowly-varying wave pressure component, while the second factor A2 represents the change of the breaking pressure component (dynamic pressure component or impulsive pressure component). The third factor A3 represents the changes in the uplift pressure. In the tbllowing chapters, the modification factors are explained for other types of caisson breakwaters. The wave height and the wavelength applied in the calculation of design wave tbrces are those of the highest wave in the design sea state. Its height is taken as H D = H ...... =

~'-P, ~-~--B~ sea side

~ .... ~

it

L

,

h,

h' _ _

~

harbour side

J_

o.oy

V

S.W.L

N

Fig. 16. Distribution o f wave pressure on an upright section.

K. Tanimoto, S. Takahashi / Coastal Engineering 22 (1994) 57-77

67

1.8H~/3 seaward of the surf zone, and as the largest wave height of random breaking waves at the water depth hh within the surf zone. The wavelength of the highest wave at the water depth h is taken as that corresponding to the significant wave period Tt/3.

H I/25 o =

3.3. Stabilio' examination of upright section The design of the upright section must be stable against sliding and overturning. To accomplish this, safety factors against sliding and overturning must not be less than 1.2. In most cases, sliding is more severe than overturning because of the relatively low crest of breakwaters in Japan. The safety factor, SF, against sliding under wave action is defined as follows:

SF~ = ~ ( W o - U)/P

(il)

in which p, denotes the coefficient of friction between the upright section and the rubble mound, Wo the weight of the upright section per unit extension in still water, U the total uplift force per unit extension, and P the total horizontal wave force per unit extension calculated by Eqs. (1) to (10). The coefficient of friction (static friction) between a concrete slab and rubble stones is usually taken as 0.6. Goda (1985) examined the stability of prototype breakwaters by using the safety factor against sliding of the upright section. Fig. 17 shows the results in a different way from the original presentation. In this figure, the abscissa indicates the safety factor against sliding calculated by the Goda formula, and the ordinate indicates the root mean square values of sliding distances observed for the upright sections under similar conditions. Analysis of the data of the prototype breakwaters obtained before 1973 revealed that no sliding occurred in any of prototype breakwaters, when the safety factor according to the Goda formula exceeded 1.2.

3.4. Bearing capaci~ of rubble mound foundation The bearing capacity of the rubble mound and the sub-soil under inclined and eccentric loads due to the weight of the upright section and wave forces must be investigated for I0

l

I

I

I

l

I

I

0 S=O 0
8

•

1.0< S

4

2 0

0

0.2

0.4

0.6

0.8

1.0

1.2

1.4

1.6

safety factor against sliding Fig. 17. Examination of the safety factor of prototype breakwaters against sliding.

68

K. Tanimoto. S. Takatu vI /CoaYt~ I E gil eeri g 22 (1994) 57 77

stability. Previously, the Japanese technical standard had employed a complex procedure to evaluate the stability of gravity structure on rubble mounds. The procedure included calculation of the maximum contact pressure, or toe pressure, which had usually been taken at 40 to 50 tf/m 2 (400 to 500 kN/m2). Kobayashi et al. (1987) proposed a new calculation method for the bearing capacity of a gravity structure on a rubble mound with the simplified Bishop method of circular slip failure analysis, which was included in the 1989 edition of the technical standards. In this procedure, the apparent cohesion is introduced for rubbles, based on the results of large triaxial tests. The standard values, the apparent cohesion C = 2 tf/m 2 and the angle of shear resistance qb= 35 °, are applied for normal rubble which is widely used in Japanese harbour construction. It is common practice in Japan to evaluate the internal angle of friction @ of sand from the N value obtained from the standard penetration test. The following standard value of is applied depending on N value: @= 40 ° for sand with N < 10 @= 45 ° for sand with N > 10 The safety factor according to the simplified Bishop method must not be less than l for breakwater subjected to wave action.

t,,rmo,,r f

sson j

Foot Protection Block

Fig. 18. Armor layer of rubble mound foundation.

,oo

K

~O 'W';, . . . . . . ' E eL Je

em 50

eF

E°

•H

J'"~P

eB sI

'

' 'r',

C

B : Wove Angle •G

W: Weight of Stones We: Colculoted Weight IN

o

'

o.o',

,

,

,I

....

J

0.05 o.l

,

,

M,~ ....

0.5 W/

,_A

l

,

,o

....

3

~ ~,o

Wc

Fig. 19. Examination of the stability of prototype armor stones.

K. Tanimoto. S. Takahashi/ Coastal Engineering 22 (1994) 57-77

69

3.5. Stability of armor units for the rubble mound foundation The stability of the armor units for rubble mounds against wave action must be investigated. The stable weight of armor units W can be expressed by the following relation: W = wrH3/3 /

{N~(

Wr/Wo - - 1) } 3

(12)

in which Wr denotes the specific weight of the armor unit, H~/3 the design significant wave height, and N~ the stability coefficient. The stability coefficient N, depends on such variables as shape of armor unit, manner of placing, shape of rubble mound foundation, wave conditions (height, period, direction) and so on. Tanimoto et al. (1982) proposed a formula to calculate the stability coefficient for two layers of quarry stones based on analytical considerations and the results of random wave experiments. Takahashi et al. (1990b) modified Tanimoto's formula so it can be applied to obliquely incident waves. That is, Ns =max{ 1.8, 1.3a+ 1 . S e x p [ - 1.5a(l - k ) ] }

(13)

in which, or= {( 1 -- K)IKI/3}(h'/Hj/3)

(14)

Ki = (4rrh'/L')/sinh(47rh'/L')

(15)

(K2) B =

max{ ce~sin2flcos2 [ (2rrx/L')cos/3], cos2/3sin 2[ (27rBM/L')cos/3] }

L' denotes the wavelength corresponding to the significant wave period at the depth h', x the distance from the wall ~
4. Impulsive forces by breaking waves 4.1. An example of breakwater damage due to impulsive wave forces In the winter of 1973/74, a typical sliding failure of a caisson breakwater occurred at Sakata Port. Fig. 20 shows a cross-section before the damage, and Fig. 21 shows sliding distances of individual caissons based on the measurements taken after storms. The breakwater extension from caisson No. 1 to caisson No. 39 is almost parallel to the depth contours and is connected to another breakwater extending seaward. All caissons with the exception

K. Tanimoto. S, Takahashi I Coastal Engineering 22 (1994) 57-77

70

IZO ~ . j

--

lOt

Tetrapod / \

- - 3 4 3 4.5_11 : ~--~

..,

~

_

_

II ~1

-,.,,

_

~

I ~4 5 1 ~~- ~ 1

14.5

,

-5

,.

3

...... L-~o_=C_~L~

Fig. 20. Cross-section of caisson breakwater before damage at Sakata Port. No. of caisson /11213141 sis 1718 I§ h oh lh 2h 311dl'd116117118h9120~21~22123~2L~25i2 6~2'7~28~29~30~31 b213313z43513~768139 I

o~ .... - ~ . U , - :

;-:--~-=----

g so

.

I00 150

:6

"-

1

" z - " ~'= =-: = -' ' z -~= > -S =

.

.

-~,

.

~"

~"

'"

:

"

~

c

""

..... '! "" date of survey

,'¢

,.-l~tT~

,.,

- / "~" \ ~" ""

/"

200 I

~--~

\ ""

~ A u g , 29,1973"

:~ 250

----Oc t, 27,1973

/

to

,..

300 I 350

,oo I

~

/

---Nov. 9,1973 ---Dec.l 2,1973 .... Jan.21,1974 -*-Feb. 6,1974

" ~ "

,

lllllll

~-

I I

Fig. 21. S l i d i n g d i s t a n c e o f c a i s s o n s at Sakata Port ( 1 9 7 3 - 1 9 7 4 ) .

of three units around the connection part slid by winter storm waves, although the final sliding distance did not exceed 3.8 m. Waves were recorded at a depth of 14 m. The maximum significant wave height during the winter was 7.2 m on 17 November. The depth at the breakwater site was 9.5 m, including the tide, which means that breaking waves acted on the breakwater during heavy storms. Model experiments with a scale of 1/25 were carried out in a wave channel for the damaged breakwater (Tanimoto et al., 1981 ). The experimental results showed that large impulsive pressures acted on the caisson, causing it to slide significantly when H = 6.88 m and T = 14 s. 4.2. Variation o f wave f o r c e s due to m o u n d configuration

Although the Goda formula can successfully be applied to the design of the upright section in most cases, it fails to predict the wave forces under certain conditions, such as when the wave forces act on the upright section with a large and high rubble mound

K. Tanimoto, S. Takahashi I Coastal Engineering 22 (1994) 5 7 - 7 7

71

foundation. For example, Fig. 22 shows the variation in the wave pressure intensity t5 averaged over the total height of the upright section. In the figure, the abscissa represents the berm width of the rubble mound foundation relative to the water depth, and the ordinate represents the thickness of the rubble mound foundation, including the armor layer, relative to the water depth. The data o f p are obtained on the basis of sliding tests of an upright section for different sizes of the rubble mound foundation by using the following relation (Tanimoto et al., 1981):

ill= IzWoJ ( 1 + I~U~/ Po)

(16)

where I is the total height of the upright section ( = hc + h' ), Wo~ is the threshold weight in water against sliding determined experimentally for a given wave condition, and Uc and Pc are the total uplift force and the total horizontal wave force calculated by the Goda formula, respectively. Therefore, p is not the actual wave pressure but is an equivalent static wave pressure intensity which allows the stability examination against sliding to be calculated from Eq. ( 11 ). In the figure, the variation of the pressure intensity (15/wHo) is shown by the isolines for the wave condition where h/L = 0.0712 and H/h = 0.719. This example illustrates how greatly wave pressure intensity is influenced not only by the height of the rubble mound foundation but also by the berm width, and that the breaking wave forces become very large when the rubble mound foundation is too high and wide. Under these wave conditions, the horizontal wave pressure (equivalent static pressure) exceeds 2.5 w~M and takes the maximum when ( h - d ) / h = 0 . 6 and BM/h = 1.5. Fig. 22 is one of the figures obtained from comprehensive sliding tests conducted in a wave tank. Since the influence of the berm width is not considered in the Goda formula, and since the non-dimensional pressure intensity predicted by the Goda formula never exceeds 2.1, the inability of the Goda formula to predict wave forces under all conditions becomes apparent. Regardless, the Goda formula is very useful as a design wave force formula, because the breakwater sections should not be designed to receive excessively large wave I0

~.~__.

I

0.9

. . . . . __ h/L=ODTI2, H/h=O.719

0 1.0213

0.8

h-d X

0.7

o6 0.5 0.4 0.3 0.2 0.1 0

0

0.5

1.0

1.5

2.0

2.5 3.0 8w/h

5.5

4.0

4.5

5.0

Fig. 22. Example of wave force variation due to mound configuration.

K. Tanimoto, S. Takahashi / Coastal Engineering 22 (1994) 57-77

72

forces in terms of the stability and construction cost of the breakwater. In fact, a low rubble mound foundation (d/h>0.6) is strongly recommended when breaking waves act on breakwaters with normal incidence.

4.3. Impulsive pressure coefficient Recently, Takahashi et al. (1992) proposed an impulsive pressure coefficient which is obtained by a re-analysis of the results of comprehensive sliding tests mentioned above. The impulsive pressure coefficient is a non-dimensional value representing the impulsive pressure component, which should be regarded as an additional effect to the slowly varying pressure. The effect of the dynamic (impulsive) pressure indicated by the coefficient o~2 in Goda's formula does not accurately estimate the effective pressure (equivalent static pressure) due to impulsive pressure under all conditions. A new impulsive pressure coefficient a~ is introduced into the Goda pressure formula. The pressure p~ at the water surface in the Goda pressure formula Eq. (2) is replaced by p, =0.5( 1 +cos/3)(h, a, +h~cr*cos2fl)wJ-1D

(17)

where a* represents the coefficients of dynamic (impulsive) pressure, a, is pressure coefficient for slowly varying pressure,/3 is wave direction, and H D is the design wave height. The correction factors A ,, A 2 and A3 are all unity. The coefficient o~* is newly expressed by a * = max{cr2,a, }

(18)

Fig. 23 is a calculation diagram for the impulsive pressure coefficient a~ obtained from sliding tests. The coefficient a~ is expressed by the product of a,~ and crib. where air ~ represents the effect of wave height on the mound, i.e..

ato = H / d H <~2d ) =2 H>2d

(19)

and O~l~represents the effect of the mound shape and is shown by the contour lines. This term is also evaluated using all = cos32/cosh61

62 ~ 0 }

= I/[cosh61(cosh62) I/2] 6~>0,, 6j=2061~ =15611 62 =4.9622

=36~2

611~0"~ 611>0,,

(21)

622 <~0 622 > 0 , ,

6~ =0.93(BM/L-O. 12)+O.36[(h-d)/h-0.6] -0.36(BM/L-O. 12) + 0 . 9 3 [ ( h - d ) / h - 0 . 6 ) ]

6.2

(20)

(22)

"~

/

(23)

The value of tr[ reaches a maximum of 2 at BM/L = O. 12, d/h = 0.4, and H/d> 2. When d/h > 0.7, cr[ is always close to zero and smaller than o~. It should be noted that the impulsive pressure decreases significantly when the angle of incidence is oblique. An

K. Tanimoto, S. Takahashi/ Coastal Engineering 22 (1994) 57-77

I

02

ct I

( alt = )

H

H<2d

L2

H>2d

t

0

7

I 0

BM_

Clio ~ a l l

:

I 0

I

I

73

/~ 0 /

jo: 10 8

l 0.2

BM L

03

04

Fig. 23. Calculation diagram of impulsive pressure coefficient. experimental result in a wave basin shows that the impulsive pressure can be neglected when the incident wave angle is above 30 ° .

5. Countermeasures against strong impulsive pressure Strong impulsive wave forces are exerted on vertical walls when the slope of sea bottom is steep and/or when the rubble mound foundation is high and wide, and when breaking waves act on them with nearly normal incidence (wave angle is nearly zero). The impulsive pressure coefficient can reflects the level of the danger associated with impulsive pressures. It is very important to avoid these conditions in the design of vertical breakwaters. 5.1. Caisson covered with wave-dissipating blocks

An alternative design to avoid strong impulsive wave forces is to use wave dissipating concrete blocks in front of the upright section as in Fig. 10. Wave forces acting on the upright section are reduced by placing wave-dissipating concrete blocks in front of the caisson. In this way, strong impulsive breaking wave forces can be minimized. Fig. 24 demonstrates how the wave pressure intensities (equivalent static pressure) differ with and without wave-dissipating concrete blocks. The pressure intensity is greatly decreased by the addition of the blocks when the rubble mound foundation is high ( d / h = 0.405). The wave forces acting on the upright section with wave dissipating concrete blocks

74

K. Tanimoto, S. Takahashi I Coastal Et gineering 22 (1994) 57-77 2.5

without blocks d/h

---*- 0 . 4 0 5 .-c-- 0 . 6 4 3 2D with blocks

d/h

WoH 1.5

0.405 --~-- 0.643

/ \

1.0

,,P" q 0.5

o

0

0.2

0.4

0.6

H/h

0.8

I.O

Fig. 24. Change of impulsive wave force by covering with wave dissipating concrete blocks.

//~t

I -

-~........ --J~Cres .

.."

t IIa

b

@

.

.

.

.

.

.-"

.

.

.

.

.

.

.

gh ff

gh~

Fig. 25. Phases for pressure calculations of perforated wall caissons.

....

_.:'_'_ . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .

Fig. 26. Wave pressure distribution at Crest-lib for perforated wall caisson.

75

K. Tanimoto, S. Takahashi / Coastal Engineering 22 (1994) 57-77 TABLE 1 Modification factors for vertical slit wall caisson Crest -I

Slit wall

As~ As2

Front wall

Crest-lla

0.85 J" 0.4 ( 0.3/a*

ALt hL2

(a* ~<0.75) (a*>0.75) 1.0

.J" 0.4 0.2/a*

Wave chamber ARI rear wall*

(a* <0.5) (a* >0.5)

201/3L'

0

0.7

0.3

0

0

0.75

0.65

0

0

(I/L'<~O.15)

{

1.4

1.6-2H/h

AR2

0

Wave chamber )tMi bottom slab

0

Uplift

Crest-lib

1.0

(I/L' >0.15)

0.56 0.5/a*

(a* ~<25/28) (a*>25/28)

1.0

(l/L'>O.15)

1.0

(H/h<~O,I) (0,1 ~0.3) 0

1.6- 2H/h 1.0

(0.1 < H / h <0.3)

(H/h>~0.3)

/~M2

0

0

0

hu ( = h3)

1.0

0,75

0.65

In the calculation of a* for the rear wall, a~ should be replaced by ot'~which is obtained with the parameters d', L' and B'M instead of d, L and BM respectively, where d' is the depth in the wave chamber. L' is the wave length at water depth d, B'M = I-- ( d - d' ), and I is the width of the wave chamber including the thickness of the perforated vertical wall.

placed in front can be calculated by the generalized Goda formula incorporating the following modifications (Tanimoto et al., 1984; Takahashi et al., 1990a): A l = A 3 = 1.0

H/h<~0.3

]

1.2-2/3(H/h) 0.3 ~0.6

A2 = 0

(24) (25)

5.2. Perforated wall caissons Another alternative to avoid strong impulsive wave forces on the upright section is to adopt the wave dissipating caisson which employs a perforated front wall and a wave chamber. Perforated wall caissons are becoming the design of choice in the construction of seawalls and breakwaters in Japan. This is because perforated wall caissons have low wave reflection and overtopping characteristics, and are highly stable due to their ability to absorb wave energy. This type of breakwater was initially intended for use in relatively calm seas, but has gradually been used in heavier, open seas.

76

K. Tanimoto, S. T.kahashi / OJa.s'tal Engineering 22 (/994) 57~77

In the design of perforated wall caissons, the pressure distributions at several important phases should be evaluated. This is because the forces on the members of caisson reach their peaks at different phases, and the peak of sliding or overturning forces does not necessarily occur when the wave crest is just in front of the caisson. Appropriate design of the members of the perforated wall caisson is vital if the caisson is to be used as a breakwater in rough seas, especially when subjected to highly impulsive waves. There are six wave phases: Crest-I, Crest-IIa and Crest-IIb, Trough-I, Trough-II and Trough-llI, as shown in Fig. 25. At Crest-I, the force on the front wall is greatest. The force on the rear wall in the wave chamber has one impulsive peak at Crest-lla and one gentle peak at Crest-Ilb. The pressure distribution at each phase is given based on the Goda pressure formula with modification factors, A~, A2, A 3, and Eq. (17) (Tanimoto et al., 1981; Takahashi et al., 1991 ). Fig. 26 shows the distribution of the pressure at Crest-Ilb. The correction factors are determined at each phase for each member. The suffixes of A like A R indicate the positions where the factors are employed. Table I shows the modification factors, A ~, A~, andA~, for a typical perforated wall caisson, a vertical slit wall caisson. As expressed by the modification factors, not only the impulsive pressure component but also the slowly-varying pressure component is reduced fbr perforated wall caissons. Most of the wave pressure formulas of new caisson breakwaters were adopted from the Goda formula as with the vertical slit caissons ( Tanimoto et al.. 1987 ).

6. Concluding remarks The design techniques of vertical breakwaters in Japan have developed as a result of abundant experiences in the field and applied research. The most important force to be considered in the design is the wave force acting on the upright section. The basic and modified Goda formula can be applied successfully to calculate the design wave forces for most conditions of vertical breakwaters. In a limited number of conditions, however, excessive impulsive forces due to the action of breaking waves are exerted on the upright section, These conditions must be avoided in the design of vertical breakwaters. The adoption of the vertical breakwater utilizing wavedissipating concrete blocks or the wave-dissipating caisson breakwaters are appropriate as alternative designs. All new caisson breakwaters described in the present paper avoid strong impulsive wave forces due to the wave-dissipating mechanism and/or the wave phase dispersion mechanism. They are no more the breakwaters just struggling against waves but the breakwaters harmonizing with waves. The examinations of the bearing capacity of rubble mound foundation and the stability of the armor units under the action of waves are also important in the design of vertical breakwaters. Recent developments in this area are certain to improve overall safety and durability. An exciting subject related with the vertical breakwater design, currently being investigated, is an assessment of stability in a probabilistic way. Takayama and Fujii (1991) proposed a method to estimate the degree of instability of the upright section which is represented by the value of probability of a sliding failure occurring during the lifetime of

K. Tanimoto, S. Takahashi / Coastal Engineering 22 (1994) 57-77

77

the breakwater. Furthermore, a method for predicting breakwater deformation, such as the sliding distance, which takes into account the statistical variability is soon to be explored.

References Danel, P., 1953. Tetrapods, In: Proc. 4th Conf. Coastal Eng., Chicago, 1L. ASCE, New York, pp. 390-398. Goda, Y., 1973. A new method of wave pressure calculation for the design of composite breakwater. Rep. Port Harbour Res. Inst,, 12 ( 3 ): 31-70 ( in Japanese) or Proc. 14th Conf. Coastal Eng., 1974, Copenhagen. ASCE, New York, pp. 1702-1720. Goda, Y., 1985. Random Seas and Design of Maritime Structures. Univ. Tokyo Press, Tokyo. Jarlan, G.E., 1961. A perforated vertical breakwater. Dock Harbour Auth., 41 (488) 394-398. Kobayashi, M., Terashi, M. and Takahashi, K., 1987. Bearing capacity of a rubble mound supporting a gravity structure. Rep. Port Harbour Res. Inst., 26(5 ): 215-252. Okada, Y., Watanabe, T., Sugawara, T. and Tanimoto, K., 1990. Recent developments of new type breakwaters in Japan. In: PIANC 27th International Congress, Osaka. Port and Harbour Bureau, Ministry of Transport, 1989. Technical Standards |br Port and Harbour Facilities with Expository Comments. Japan Port and Harbour Association (in Japanese), or Technical Standards for Port and Harbour Facilities in Japan, new edition 1991. The Overseas Coastal Area Development Institute of Japan, Tokyo, Takahashi, S., 1988. Hydrodynamic characteristics of wave-power extracting caisson breakwater. In: Proc. 21st Conf. Coastal Eng. Conf., Malaga. ASCE, New York, pp. 2489-2503. Takahashi, S., Tanimoto, K. and Shimosako, K., 1990a. Wave and block forces on a caisson covered with wave dissipating blocks. Rep. Port Harbour Res. Inst., 29( I ) : 53-75 ( in Japanese ). Takahasbi, S., Kimura, K. and Tanimoto, K., 1990b. Stability of armor units of composite breakwater mound against oblique waves, Rep. Port Harbour Res. Inst., 29(2) : 3-36 ( in Japanese ). Takahashi, S., Shimosako, K. and Sasaki, H., 1991. Experimental study on wave forces acting on perforated wall caisson breakwaters. Rep. Port Harbour Res. Inst., 30(4): 3-34 (in Japanese). Takahashi. S., Tanimoto. K., and Shimosako, S. 1992, Experimental study of impulsive pressures on composite breakwaters. Rep. Port Harbour Res. Inst., 31 ( 5 ) : 35-74. Takayama, T. and Fujii, H., 1991. Probabilistic estimation of stability of slide for caisson type breakwaters. Rep. Port Harbour Res. Inst., 30( 4): 35-64 (in Japanese ). Tanimoto K. and Goda. Y., 1992. Historical development of breakwater structures in the world. In: Institution of Civil Engineers (Editors), Coastal Structures and Breakwaters. Thomas Telford, London, pp. 193-206. Tanimoto K., Takahashi, S. and Kitatani, T., 1981. Experimental study of impact breaking wave forces on a vertical-wall caisson of composite breakwater. Rep. Port Harbour Res. Inst., 20( 2 ) : 3-39 ( in Japanese ). Tanimoto K., Yagyu, T. and Goda, Y., 1982. Irregular wave tests for composite breakwater foundation. In: Proc. 18th Conf. Coastal Eng., Capetown. ASCE, New York, pp. 2144-2163. Tanimoto K., Takahashi, S. and Myose, K., 1984. Experimental study of random wave torces on upright sections of breakwaters. Rep. Port Harbour Res. Inst., 23(3): 47-99 (in Japanese). Tanimoto, K., Takahashi, S. and Kimura, K. 1987. Structures and hydraulic characteristics of breakwaters - - The state of the art of breakwater design in Japan. Rep. Port Harbour Res. Inst., 26( 5 ) : 11-56. Tanimoto K., Endoh, H. and Takahashi, S., 1992. Field experiments on a dual cylindrical caisson breakwater. In: Proc. 23rd Conf. Coastal Eng., Venice. ASCE, New York, pp. 1625-1638.