- Email: [email protected]
Aftershock response of RC buildings in Santiago, Chile, succeeding the magnitude 8.8 Maule earthquake
Engineering Structures 76 (2014) 324–338
Contents lists available at ScienceDirect
Engineering Structures journal homepage: www.elsevier.com/locate/engstruct
Aftershock response of RC buildings in Santiago, Chile, succeeding the magnitude 8.8 Maule earthquake Anne Lemnitzer a,⇑, Leonardo M. Massone b, Derek A. Skolnik c, Juan C. de la Llera Martin d, John W. Wallace e a
University of CA, Irvine, Dept. of Civil & Environmental Engineering, 4149 Eng. Gateway, Irvine, CA 92697, USA University of Chile, Santiago, Blanco Encalada 2002, Santiago, Chile Kinemetrics Inc., 222 Vista Avenue, Pasadena 91107, CA, USA d Department of Structural and Geotechnical Engineering, Pontificia Universidad Catolica de Chile, and National Research Center for Integrated Natural Disaster Management CONICYT/FONDAP/15110017, Vicuna Mackenna 4860, Santiago, Chile e University of California Los Angeles, 405 Hilgard Ave, Los Angeles 90024, USA b c
a r t i c l e
i n f o
Article history: Received 10 November 2013 Revised 1 July 2014 Accepted 2 July 2014
Keywords: Seismic monitoring Reinforced concrete structures Instrumentation
a b s t r a c t Between March 13th and 28th 2010, a team of U.S. researchers, professionals, and local collaborators instrumented four reinforced concrete (RC) buildings in Santiago, Chile, to measure aftershock response data following the February 27, 2010 MW 8.8 earthquake. The selected buildings, designed according to NCh433.Of96 (similar to ACI 318-95), represent typical construction, e.g. moderate to high rise office buildings with large office space and inner core shear walls and mid-rise residential shear wall buildings. Two of the instrumented buildings were undamaged, whereas one building suffered only minor nonstructural damage and the fourth building exhibited more significant structural damage such as column buckling and shear wall cracking. Instrumentation consisted mainly of uni-axial and tri-axial accelerometers as well as some displacement transducers. Records for several aftershocks were captured during a period of one month. Collected data were processed and system identification algorithms were used to determine dynamic building- and modeling parameters (e.g. building periods, acceleration amplification, inter-story drifts, building rocking and torsion, and assessment of diaphragm in-plane rigidity). A comparison between linear elastic models used in engineering practice and measured data was performed for one building. Information related to the instrumented buildings, data collected, and analytical results is presented, along with practical lessons learned conducting monitoring studies in the aftermath of this very strong earthquake. Ó 2014 Elsevier Ltd. All rights reserved.
1. Introduction Building instrumentation has become essential in seismically active areas to gain insight into building behaviour under earthquake loading. For example, the city of Los Angeles has included modest levels of instrumentation in the applicable building code since 1965. Since 2010, more extensive levels of instrumentation are required for tall buildings designed using non-linear response history procedures of ASCE 7 Chapter 16 [1,2]. The California Strong-Motion Instrumentation Program [3] and the Advanced National Seismic System [4] are tasked with operating larger instrumentation programs for the state of California and the Federal level respectively. A detailed history of building
⇑ Corresponding author. Tel.: +1 (310)986 5255. E-mail address: [email protected] (A. Lemnitzer). http://dx.doi.org/10.1016/j.engstruct.2014.07.003 0141-0296/Ó 2014 Elsevier Ltd. All rights reserved.
instrumentation programs within the United States is available in Celebi [5]. Data from permanently and temporarily instrumented buildings have enabled important studies in system identification and model updating [6,7] damage detection and long-term health monitoring [8–11] and have led to important code provisions such as fundamental period formulas [12,13]. The goal of these programs and provisions is to monitor various building construction types to gain better understanding of building performance under dynamic loading and enable the improvement of current design codes and practices. Damage and failures observed after the MW 6.7 Northridge earthquake in 1994 led to significant changes to buildings codes; however, since 1994, the United States has not experienced a large magnitude earthquake (>7.0); therefore, major changes over the past decade or so have relied on observations from, and studies of, earthquakes occurring outside of the US. While permanent instrumentation programs (such as those
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
previously described) are gaining traction around the world, their evolution is slow due to the high costs associated with equipment installation and maintenance, as well as the lack of building instrumentation in locations that have experienced strong shaking. In turn, rapidly deployable temporary instrumentation programs in areas with frequent seismic events provide a unique opportunity to collect significant amounts of data over a short period of time. The NEES@UCLA facility includes equipment and expertise to enable rapid system deployments to enable rapid aftershock monitoring. Aftershock studies, in particular, offer excellent possibilities to capture exclusive data for buildings damaged during the main shock, such as the validation of modeling assumptions during the design process or evaluation of various retrofit schemes. Several hundred aftershocks followed the main rupture of over 550 km long on February 27th, 2010s MW 8.8 Maule earthquake in Chile. Most large magnitude shocks (M > 6.0) occurred within 24 h after the main event [14] while moderate aftershocks continued for over 5 weeks. The epicenters of the majority of aftershocks were located near the epicenter of the Mw 8.8 earthquake (i.e. near Concepcion), but were felt in radial distances of up to 300 km, including the areas of Vina del Mar/Valparaiso, Santiago and Mendoza. A team of researchers and professionals visited several reinforced concrete buildings in Chile’s capital Santiago as part of a reconnaissance effort in March 2010 and selected four buildings for temporary instrumentation over a period of 4–8 weeks. The buildings were chosen based on their availability/accessibility and design similarities to reinforced concrete structures present in the United States in order to make the measured data most beneficial for both countries. The selected structures were designed using the local, pre-MW 8.8 design recommendations for reinforced concrete buildings, which were taken from the ACI 318-95 [15] specifications in the United States, with the exception of the boundary element confinement provisions for shear walls. Chile’s design code [16] was revised between 2010 and 2011 and now includes many lessons learned from the 2010 Maule Earthquake and its reconnaissance efforts such as modifications of wall boundary transversal reinforcement, axial load limitations, damage limitations, minimum thickness requirements of confined walls. A summary of the adapted changes along with a thorough explaining of the reasoning can be found in Massone and Massone et al. [17,18]. The objective of this study was to collect structural performance data that would enable a basic analysis to quickly provide the most important building dynamic and modeling parameters such as periods, acceleration amplification factors, inter-story drifts, and rocking responses as well as torsional effects and damping ratios (if possible). It also forms the foundation to guide subsequent in depth studies such as nonlinear structural modeling and comprehensive system identification analyses. This study also forms an initiative to create a unique database of structural input and response data from aftershocks and main shocks. It represents one of the first efforts to organize a spontaneous team to visit a foreign country immediately following a seismic event that imports of all
325
instrumentation and data acquisition technologies from the United States. Valuable practical lessons were learned during this effort and applications and limitations with monitoring studies in foreign countries are shared with the reader and are used to enhance the success of future similar efforts. All data of this instrumentation effort are publicly available at the NEES project warehouse [19–22] which can be found at https://nees.org/warehouse/ report/project/941. 2. Selected case studies The four buildings selected for instrumentation are reinforced concrete structures located in downtown Santiago and are depicted in Fig. 1. Buildings 1 and 2 are situated in Santiago’s business district ‘‘Las Condes’’ in the North-East of the city. Building 3 is located in a prominent residential district ‘‘Vitacura’’ and Building 4 was located in Santiago’s Huechuraba district, an area known for very soft soil conditions. The following paragraphs will provide a brief overview of the building’s structural systems along with observations of successful performance or damage during the main shock of the February 27th, 2010 earthquake. Particular focus is paid to Buildings 2 and 4, since most measurements and analyses were conducted for those structures. Buildings 1 and 4, are explained in brief; the reader will be referred to the NEES project warehouse for further information and ambient vibration data. Building 1 (Fig. 1) is a 25 story high-rise office building, with additional five subterranean parking levels, whose lateral force resisting system consists of two inner-core shear wall elevator shafts complemented by an external beam-column frame system. No earthquake damage (neither structural nor architectural) was observed in this building (Fig. 2). Building 1 was instrumented with four tri-axial accelerometers on the four exterior corners of the ground floor and four corresponding accelerometers on the roof (25th floor). Instrumentation was installed over a period of four days, measuring four aftershocks with rather small peak accelerations between 0.6 cm/s2 and 2.25 cm/s2 (0.057–0.23%g). Building 1 was the first building instrumented and challenged the instrumentation team with many issues related to the equipment installation such as the continuation and interference of business operations, scouting most suitable locations for sensors and cable runs (logistically and structurally), running cables over the entire building height of (84.15 m/25 stories), and introduction of signal noise stemming from long cables, logistical aspects and equipment purchase on site (battery powered equipment and overnight recharging, outlet power converters). Several hours of ambient vibration, as well as two small aftershocks, were captured in Building 1 and are stored on the NEES project warehouse website along with photographs and instrumentation plans. Building 2 (shown in Fig. 1) is a 10 story reinforced concrete building with six levels of subterranean parking. The building was under interior construction at the time of instrumentation and provided several unoccupied floors for easy access and installation of
Fig. 1. Instrumented Building 1–4 (left to right) in Santiago, Chile.
326
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
Fig. 2. Building 1, no damage detected on façade or soft story.
sensors. The structural system consisted of small inner core shear walls formed by the elevator shaft and staircase (thickness ranging from 20 to 30 cm) and an external beam-column framing system with 0.82 0.4 m columns in the long direction and 0.5 m 0.4 m columns in the short direction spaced at 2.5 m and 3.6 m center to center in each direction respectively. Column dimensions on the 10th floor reduced to 0.5 m 0.4 m in the long direction. Typical column reinforcement consisted of A630-420H steel (fy = 420 MPa), #8–10 bars around the column periphery and transverse reinforcement of #3–4 hoops spaced at d/2. The total floor area of Building 2 is approximately 3209 m2. The office
building is a representation of a structure in the Occupancy Category 2 [23]. Similar to Building 1; the structure of this building did not suffer any structural damage during the Mw 8.8 earthquake or its subsequent aftershocks and represents a success story for well-designed RC buildings. Fig. 3a and b shows the location of Building 2 in the business district of Las Condes, as well as an interior view of the unoccupied office space on floor 10. Building 3 (Fig. 1) was located in a newly developed luxurious residential district of Santiago in the north-east part of the city and represents a 10-story reinforced concrete shear wall building with one level of underground parking. Fig. 4(a)–(c) shows the
Fig. 3. (a and b) Location of Building 2 (left) and view of interior office space used for instrumentation.
Fig. 4. (a–c) Building 3 front view (a), side view (b) and damage at 1 story level (c).
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
327
Fig. 5. First floor plan view with more severely damaged elements.
building in front and side view. Fig. 5 presents a structural plan of its first floor. Floors 3–8 were typical floors and maintained many of the configurations shown in the first floor. At the time of deployment (March 2010) only floors 2 and 9 were unoccupied and chosen for instrumentation. Aside from floors 2 and 9, additional sensors were placed at the subterranean level and the roof. A total of 10 accelerometers and four linear voltage differential transducers (LVDTs) were installed in the building. The building suffered moderate structural damage at the subterranean parking level which consisted mainly of spalling of the concrete cover, section cracking and longitudinal rebar buckling at the column and wall to ceiling interface (see Fig. 6). Damage on the ground level included column failure in two structural columns supporting the second floor (Figs. 5 and 7) which
was introduced through large axial compression forces. Similar instabilities were observed by Wallace et al. [24] for other RC structures in Chile. Significant shear cracking was noticed on interior and exterior shear walls (Fig. 8) Smaller shear cracks were detected in floors 2–10 and reoccurred consistently in exterior walls when window (or door) openings were discontinued in floors located above or below (i.e., producing an irregular opening pattern over the building height). The structure is a commonly used type for residential buildings in Chile. Thicknesses of shear walls and slabs were generally 17 cm and 15 cm, respectively. Typical material properties are a concrete strength of H30 (f0 c = 25 MPa) and reinforcing steel with Grade A630-420H (equivalent to US Grade 60, fy = 420 MPa). The building was designed according to NCh433.Of96, following most of the ACI 318-95 requirements
Fig. 6. (a and b) Moderate damage at underground parking level Building 3.
328
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
were measured but no aftershocks were captured in this building during the instrumented period. This data, along with structural plans, is available on the project warehouse for future studies and may be particularly beneficial during investigations in future seismic events, in particular for soil structure interaction studies, given the soft soils at the site of Building 4. The remainder of this manuscript will limit its focus on Buildings 2 and 3 for which the most aftershocks with quality data were obtained. Both buildings incorporate typical lateral load resisting systems (moment frames or shear walls) and provide a good representation of the overall data collection effort and building analyses. Although measured accelerations were small, recordings were sufficient to conduct fundamental analyses of the respective buildings and provide a comprehensive insight into important building parameters. 3. Instrumentation
Fig. 7. Two column failures at ground level due to axial and shear loading.
with the exception of the boundary element confinement of shear walls. As it can be seen in Fig. 5, the area of walls was distributed almost equally in both orthogonal directions of the building. The wall to floor area ratio in each direction was about 1.5%, which is typical but could be considered in the lower end of wall to plan ratios in Chile [18]. The static loads in the structure at the time of the measurements correspond to the self-weight (dead load) including structural and non-structural elements, and certain live loads which were expected to be relatively small since the buildings was partially uninhabited during monitoring. Building 4 (Fig. 1) was a five story moment frame office building which was under interior construction (partitions, ceilings) at the time of instrumentation. Large amounts of ambient vibrations
The instrumentation in Buildings 2 and 3 consisted of uniaxial and triaxial Episensor accelerometers with a measurement range of ±4 g. Additionally, Building 3 was instrumented with several linear voltage differential transducers (LVDTs) installed at the damaged shear wall (see Fig. 9). All data were recorded with Quanterra Q330 digitizers and processed with MATLAB [25]. Data were recorded at a sampling rate of 200 Hz, and recording was triggered when accelerometer readings exceeded a threshold of 0.005 g. Each recording included 15 s of pre- and post-event data. Several hours of ambient vibration data were also recorded during day and night time to establish baseline behaviour. The monitoring system was powered using large 12 V batteries which were locally purchased and charged onsite. All monitoring equipment belonged to NEES@UCLA and was brought into Chile using a temporary instrumentation permit issued by the customs and immigration office upon arrival in Santiago. Fig. 10 shows a 3D instrumentation layout of Building 2 with the sensors indicated on floors 1 and 10. Fig. 11 shows the plan layout and exact location of the accelerometers installed on the respective floors. Floor one was instrumented with three uniaxial accelerometers in each corner of the building and floor 10 was instrumented with one triaxial accelerometer in each corner. All instrumentation was directly mounted on the reinforced concrete slabs. Four aftershocks with moment magnitudes ranging between 4.7 and 5.1 were measured in Building 2. The amplitude of translational acceleration over the multiple aftershocks ranged from 0.001 to 0.01 g. Fig. 12 shows the simplified 3D schematic of Building 3. This structure was instrumented with one triaxial accelerometer at the center of the building on the basement slab, i.e. on the concrete slab at underground parking level, three triaxial accelerometers on the second floor (installed on tile floors, or directly on concrete slab), one triaxial accelerometer at the ninth floor (on concrete
Fig. 8. (a and b) Shear wall damage occurring at discontinuity of window openings on the 1st level, outside (a) and inside view (b).
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
329
Fig. 9. Instrumentation of damaged shear wall in Building 3.
Fig. 12. Instrumentation layout in Building 3.
slab) and three uniaxial accelerometers at the roof (installed at the concrete slab). A layout of the floor plan with the instrumentation is illustrated in Fig. 13. Between March 18th and May 5th, a total of 30 aftershocks were recorded in Building 3 with peak ground horizontal accelerations ranging from 0.5 cm/s2 to 7 cm/s2 (0.05–0.7%g). A total of 22 aftershocks were carefully checked for the proceeding data evaluation and analysis. 4. Results 4.1. Aftershock recordings in Building 2
Fig. 10. Instrumentation layout in Building 2.
A M4.7 aftershock recorded on March 25, 2010 at 10:59:35 am local time (12:59:35 pm UTC time) was selected to study the seismic response of Building 2 (Fig. 14). The epicenter was located about 17 km north of Pichilemu, which is located about 200 km south of Santiago, at a depth of 34.5 km. This epicenter coincides with the region of largest energy liberation during the earthquake. Prior to processing the measured structural accelerations, the data set was adjusted via filtering with linear bandpass filters and baseline corrections to eliminate noise. Several different sets of measured accelerations at the corners were used to calculate the three dimensional (two translational
Fig. 11. Location of instrumentation in Building 2.
330
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
Fig. 13. Location of instrumentation in Building 3.
and torsional) slab motions at a central point assuming that each slab behaves as a rigid diaphragm. Results of this analysis showed that the rigid diaphragm assumption is sound. Fig. 15 describes the relationship between local translational acceleration at any given point on a floor (ui and vi corresponding to the ith sensor) and the central (geometric) story motions (u0, v0, and h0). First floor accelerations with maximum values of 1.2 cm/s2 both, in EW and NS directions were found. Floor 10 shows center accelerations of 3.5 cm/s2 and corner accelerations of 4.5 cm/s2 in the EW direction, and center accelerations of 2.8 cm/s2 and corner accelerations of 5.0 cm/s2 in the NS direction. The amplification of acceleration between floor one and ten is in the range of 2.0–4.5 (center and corner respectively). As expected through comparison of floor corner and center accelerations, torsion was apparent on the 10th floor. Fig. 16 illustrates the maximum torsional acceleration on the first and tenth floor of 1.273 104 rad/s2 and 3.386 103 rad/s2 respectively. Double numerical integration of recorded accelerations enabled the calculation of story displacements. In the E-W direction, maximum displacements of 0.017 cm and 0.104 cm were measured for floors one and ten, respectively. The total inter-story drift between floors one and ten was 0.088 cm. In the N-S direction, maximum displacements of 0.015 cm and 0.061 cm were recorded for floor one and ten respectively, resulting in an inter-story drift of 0.046 cm, which is approximately half of the drift measured in the orthogonal direction. Displacement measurements confirm minimal torsion at level one, but twisting is observed on floor 10. The maximum relative torsional displacement on the tenth floor is 4.50 105 rad. The XY displacement of the center of floor ten is shown as particle motion – time history for the duration of the aftershock in
Fig. 17. The total displacement range in the EW direction spans over 1.8 mm and the total displacement range in the NS direction is approximately 1.35 mm. Fast Fourier Transforms (FFT) (Fig. 18) were computed to estimate buildings transfer functions (Fig. 19). Transfer functions were developed for E-W, N-S and torsional direction to identify the dynamic parameters of the first and second modes of the structure. From Fig. 19, it can be seen that the first natural mode at about 0.89–0.91 Hz (T = 1.1 s) is predominantly NS but evident in both EW and torsional directions, thus it is most likely a coupled NS-torsional mode with a close second mode in the EW directions appearing at 0.96 Hz (T = 1.04 s). The identification of coupled modes is important because it means that the lateral forces developed in the structure’s resisting elements are generally higher than they would be for a typical symmetrical building with planar-only, well-spaced modal responses [26]. Such findings should be taken into consideration during construction of analytical models and interpretation of results. A possible third mode of the structure is evident in the NS direction at 2.35 Hz (T = 0.43 s). Fig. 20 shows a simple analysis using aftershock and ambient data to investigate the relationship between the frequency response of the building in EW and NS direction and the intensity of the ground movement in form of peak floor accelerations at the 10th floor of Building 2. A clear trend of building frequency reduction with increasing floor acceleration can be observed, which means that as accelerations increase, the overall building period becomes larger. This apparent nonlinearity has been affirmed in countless similar studies and is typically explained by citing the nonlinear behaviour of soil–structure interaction and the reduced contribution of non-structural components [6].
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
331
Fig. 16. Acceleration time histories for Building 2.
Fig. 14. Map of approximate aftershock locations.
Fig. 17. Relative displacement–time history for 10th floor particle motion at the center of the slab – Building 2.
Fig. 15. Terminology of displacements and rotations in @ Center & Corners.
4.2. Aftershock recordings in Building 3 A 5.8 Magnitude aftershock which occurred on May 2nd, 2010 at10:52:30 local time (14:52:39 UTC) at a depth of 32.9 km, 12 km north of Pichilemu (see Fig. 14, http://www.sismologia.cl) was identified as the strongest aftershock in terms of absolute accelerations, and selected for detailed processing in Building 3. The peak ground accelerations (PGA) reached about 4.3 cm/s2 in EW and 7.2 cm/s2 in NS direction. The structural accelerations measured in the building are presented in Fig. 21 and Table 1.
Again, a band-pass filter with pass frequencies ranging from 0.1 to 25 Hz was implemented for data pre-processing. The amplification of accelerations from ground to roof was calculated to be 5.3 in the EW direction and 3.4 in the NS direction. Fig. 22 shows the amplification of the roof peak acceleration vs. PSA for more than 20 aftershocks recorded in Building 3. There is a clear linear trend of amplification increase up approximately 5 cm/s2, after which it tends to remain constant or even decrease. Similar results were observed by Moroni et al. [27] for a permanently instrumented, four story, RC/masonry building in the Andalucia Community in Chile. Measurements during the main shock of the 2010 Maule earthquake indicated an increase in roof amplitude up to peak ground accelerations of 20 cm/s2 which a constant ratio thereafter [27].
332
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
Fig. 18. Fast Fourier Transforms for Building 2 (FFT).
Fig. 20. Building 2 – EW and NS frequencies vs. peak floor accelerations (PFA). Fig. 19. Building 2 transfer functions.
The exact location of all sensors in Building 3 was previously presented in Fig. 13. Given the larger amount of sensors installed, this building allowed for more analyses than the previous case. Selected parametric studies and simply modeling analyses are shown hereafter. Second floor accelerations in Building 3 were estimated using a least square analysis since the three triaxial sensors installed at that level (see Fig. 13), over-determined the assumed 3 translational DOF of the floor (x, y and z). Given the instrumentation configuration on the 2nd floor, torsional and translational accelerations were determined. The maximum torsional acceleration reaches about 0.004 rad/s2. The largest translational acceleration measurement in floor 2 reached about 10 cm/s2 (1%g) in both NS and EW direction. Thus, a sensor located 10 m away from the central sensor would result in an increment of translational acceleration of 4 cm/s2 (0.004 * 1000) due to torsion, which corresponds to about 40% of the translation of the central sensor, indicating that the torsional component has a significant impact in the response in Building 3.
Fig. 21 (b) shows the EW and NS absolute displacements components of all instrumented floors. The displacements were determined by double-integrating the accelerations of each level and applying a high-pass filter with 0.2 Hz low end corner frequency to overcome residual displacements. The peak displacement increased from about 0.08 cm in the NS direction and about 0.04 cm in the EW direction on the ground floor to 0.23 cm in NS and 0.21 cm in EW direction at the roof level, indicating a displacement amplification over the entire building height of 4. Displacements amplitudes at the 9th floor and the roof were found to be very similar in magnitude (Table 1). Fig. 23 shows particle motion as absolute displacements in the NS vs. EW direction measured for various levels in the building. It can be seen that floors ‘Ground’, 2 and 9 present patterns that move over the entire plane (uniformly circular). The roof level in turn shows a clear tendency to move along a diagonal path, which is consistent with the previous observation of the roof’s EW and NS components. The roof sensor, which is eccentric relative to the other sensors, is influenced by a torsional mode given the synchrony between both displacement components. Fig. 24 presents
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
333
Fig. 21. (a and b) Story accelerations (left) and displacements (right) in EW and NS direction for Building 3.
Table 1 Building 3 – floor accelerations and displacements.
Ground 2nd 9th Roof
Accelerations (cm/s2)
Displacements (cm)
EW
NS
EW
NS
4.3 8.5 8.93 22.6
7.2 13.3 14.5 24
0.05 0.07 0.18 0.21
0.08 0.11 0.2 0.23
Fig. 23. Particle motions Building 3.
Fig. 22. Roof acceleration amplification vs. PSA in EW and NS direction in Building 3.
the transfer functions of the relative displacement response vs. the normalized ground acceleration determined at each instrumented floor. In order to smooth the observed response a Welch procedure was implemented. From the analysis, the main frequency (f) and damping ratio (b) were determined to be f = 1.09 Hz and b = 3.0% in NS direction, and f = 1.2 Hz, b = 2.2% in the EW direction. The largest amplitude of the relative displacement response occurred at a frequency of 1.09 Hz in the NS direction at the roof level. The maximum absolute accelerations of selected floors were compared with pseudo spectral accelerations (PSA) determined for a single degree of freedom system (SDOF) using the measured building parameters (natural frequency and damping ratio) that were observed in the corresponding direction (i.e., NS: f = 1.09 Hz, b = 3.0%; EW: f = 1.2 Hz, b = 2.2%). The results for both directions are presented in Fig. 25. A trend line (best fit analysis) is included to show the linear trend between the maximum floor
accelerations and the intensity of the 22 aftershocks used for this analysis. Although, increases of PSA result in increases of maximum acceleration at different levels, the 2nd floor presents little correlation with PSA (R2 < 0.4). Fig. 26 shows a comparison of the maximum relative displacements of various floor levels with the results of the pseudo spectral analysis. Good correlation is observed between the displacement and PSA (R2 > 0.8) for all levels of Building 3. Additional studies were conducted for absolute displacements and showed consistent results when compared against peak ground displacements (PGD). It can also be seen that close trends are obtained for the 9th floor and roof. All curves aim for zero displacement at zero PSA. A comparison of the overall natural building frequency (3 modes) in NS direction vs. pseudo spectral accelerations is shown in Fig. 27 using different aftershocks and ambient vibration data. Pseudo spectral accelerations were determined using first mode natural periods and damping as described above. For ambient vibration cases, PSA values were assumed zero. Although there is scatter, the trend line in Fig. 27 shows a clear tendency of frequency reduction with increasing magnitude of the ground movement, which is consistent with observations made for Building 2 in Fig. 20. This trend may be associated with the stiffness variation due to opening and closing of small cracks in the structural elements of the building.
334
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
Fig. 24. Transfer functions Building 3.
Fig. 26. Comparison of relative floor displacements with pseudo spectral accelerations (PSA). Fig. 25. Comparison of measured floor accelerations with pseudo spectral accelerations (PSA).
The instrumentation in Building 3 included the installation of linear variable differential transducers (LVDTs) across a damaged shear wall on the first floor of the building. A photograph of the
wall and the sensor instrumentation was previously shown in Fig. 9. One pair of LVDTs was installed in diagonal configuration to enable the calculation of shear deformations during the aftershock. Another pair of vertical LVDTs was placed at both ends of the wall, spanning over the entire story height to determine the
335
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
Fig. 27. Building frequency vs. pseudo spectral accelerations (PSA).
rotation due to flexural deformations. Since deformations are small, the diagonal sensors allowed for estimating the shear component of the first story lateral deformations in the wall plane. The flexural component was estimated by assuming a concentration of such deformation at wall base and that overall deformations in the wall are dominated by shear. A time history of shear and flexure displacements during the March 25 aftershock is presented in Fig. 28. The results indicate that the deformations are dominated by the shear component reaching about 0.2 mm of lateral deformation, whereas the flexural component reaches about 0.04 mm, which is only 20% of the shear component. 4.3. Simple modeling studies performed in Building 3 The amount of building information and data measurements available for Building 3 enabled a basic modeling study using the structural program ETABS [28]. The building was modeled as a 3D structure based on the original construction documents. All elements were considered linear elastic, and diaphragms were considered rigid in their plane. In order to compute the stiffness of the elements, gross concrete sections were used for the analysis, but later modified to account for degradation of stiffness in order to represent the elements in the state in which the aftershock measurements took place (i.e. damaged state). The structure was assumed to be fully fixed in all degrees of freedom at the basement–ground interface.
The gravity loads were uniformly distributed across all floor levels. A superimposed dead load of 0.1 ton/m2 (excludes the self-weight of the structure) and a live load of 0.2 ton/m2 were calculated and applied in the analytical model. The total gravity load per floor acting in the structure (dead and live, including self-weight) results in a distributed weight of approximately 1.0 ton/m2. The process of construction, exposure, maintenance, ageing and loading (in particular seismic loading), contributes to the variation of material properties in a building; for this specific case: the properties of the reinforced concrete. Considering the limitations of the available information, recommendations by FEMA 356 [29] were used to incorporate reduction factors (RF) in the ETABS model. Hereby, factors that consider different levels of cracking for the respective elements were most crucial. The reduction factors are shown in Table 2. For the purposes of this analysis, all columns stiffness’ values were reduced by 50% (meaning I = 0.5Ig, where Ig, is the moment of inertia of the gross section). Walls stiffness reduction was varied and a sensitivity study for this parameter was conducted. To enable comparisons, a baseline model (Model 1 – B) was established using the standard values in Table 2. The baseline model included a live load of 0.2 ton/m2 and a dead load of 0.1 ton/m2 in addition to its regular self-weight. For the structural stiffness of the columns a reduced value of 0.5Ig was used, and for the structural stiffness of the walls a reduced value of 0.8Ig was used. The stiffness of the severely damaged elements on the first floor was set to zero. (i.e., elements highlighted in Fig. 5). In a second model (Model 2 – IL), the stiffness parameters were kept identical, but gravity loading was increased by 0.2 ton/m2 (live loads only) which resulted in a total increase of the mass of the structure by about 20%. A third model was developed in which the loading assumptions corresponded to the baseline model, but the wall stiffness was varied (Model 3 – WS). In this model, all other stiffness reductions were removed (RFs = 1.0), all vertical elements were treated as walls and a sensitivity study of the wall stiffness alone was investigated. The severely damaged shear wall and the columns shown in Fig. 5 at the first floor were considered a zero stiffness elements. All other walls were assigned reductions of 1.0Ig (called WS100), 0.8Ig (WS80), 0.6Ig (WS60) or 0.4Ig (WS40). The ETABS model was run for all three cases and subcases and modeling results were compared with measured and back-calculated structural parameters from the monitoring effort. Table 3 shows a summary of the fundamental period of Building 3 determined for the various cases via ETABS, which corresponds to oscillations in the NS direction. The experimental fundamental period determined from building instrumentation revealed a value of 0.9 s. Analytical results show that the baseline model (Model 1 – B) is in closest agreement with the measure results. The model with the modestly increased gravity loading (Model 2 – IL) also increased the period of the structure, but maintains the same level of error as the baseline model. The sensitivity study (third case studies Models 3 – WS) indicates that the error increases significantly (up to 20% compared to measurements) when no reduction in stiffness is considered (WS100). When only the wall stiffness is Table 2 Reduction factors (based on Tables 6 and 5 – FEMA 356).
Fig. 28. Shear and flexural components of total wall displacements.
Element
Reduction factors
Beams Post-tension beams Columns – compression > 0,5 AgFc0 Columns – compression < 0,3 AgFc0 Walls Crack walls Slabs
0.5 1 0.7 0.5 0.8 0.5 0.5
336
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
Table 3 Fundamental periods of analytical model cases. Model Model Model Model Model Model Model
Fundamental period (s) 1 2 3 3 3 3
– – – – – –
B IL WS100 WS80 WS60 WS40
0.89 0.93 0.72 0.88 0.95 1.07
reduced to the original FEMA value, which is 80% (WS80), the period changes to 0.88, indicating that the modification of the flexibility of other elements different than the walls is small. This result is expected given that the structure has only very few columns and beams and is modeled like a rigid diaphragm in Etabs. A further stiffness reduction to 60% showed a continued increase in the fundamental period (T for WS60 = 0.95 s). The model with stiffness reduction to 40% showed a 19% error compared to the measured results and the model period is approximately 1.07 s. Measured and analytical time histories for accelerations and displacement enabled a simple comparison of local responses within the structure. By retrieving absolute accelerations and displacements at the locations where sensors were installed in the building, data can be directly compared without further treatment and introduction of possible errors. Figs. 29 and 30 show the
absolute acceleration records for the second floor, the ninth floor and the roof of the building. For the ninth and roof level only the NS direction is shown. In general, the model results provide a good resemblance of the shape and magnitude of sensor measurements. This indicates that the fundamental period as well as local response of the structure is adequately captured. Floor two (Fig. 29) shows a nice agreement in shape and length of the acceleration record, even though amplitudes provided by the model appear slightly less than recorded data (predominantly in the NS direction). Fig. 30 presents the comparison of data and model results for the 9th floor and similar observations can be made. The measured results tend to show a slightly smoother time history while the model shows a more distinct phase-like acceleration variation. A data-analysis comparison of the roof level showed that the amplification of the acceleration at higher floor levels is also captured in the model with increasing absolute peak accelerations from about 7 cm/s2 at the 2nd floor to about 20 cm/s2 at the roof. Lateral deformation measurements recorded at the damaged wall can be compared to the predictions from the model (baseline model was used in this case). The measurements indicate that most of the deformation originates from shear, which is consistent with the failure mode. Fig. 31 shows a comparison between model results and measurements. The top figure depicts the combined lateral deformations (shear and flexure) measured during the aftershock. The middle figure shows the modeling results, for the
Fig. 29. Absolute accelerations at the 2nd floor in E-W (left) and N-S (right) direction.
Fig. 30. Absolute accelerations at the 9th floor (left) and roof (right) in N-S direction.
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
Fig. 31. Deformation of damaged wall: top: measured total lateral deformation (shear and flexure), middle: model result for lat. deformations 0% wall stiffness (damaged wall), bottom: model results for lat. deformations – 100% wall stiffness (undamaged wall).
base model (Model 1 – B) in which the stiffness of the damaged wall was taken as zero (i.e., EI = GA = 0). The bottom figure shows the lateral deformation observed in the model when the wall was modeled with 100% of its stiffness. Results indicate that the in-situ measurements show smaller deformations than the model (Model 1 – B) predictions. This difference could be associated with the precision of the measurement equipment for such small deformation levels. A comparison between the two model results (one with 0% wall stiffness and one with 100% wall stiffness) indicated that there is a reduction in the deformation prediction, but the differences are rather modest compared to the significant change in the wall stiffness assumption. This result is expected since a rather large amount of walls were present in the structure. The displacement of the damaged wall is governed by all other connecting elements, such as other walls and slabs. 5. Lessons learned The first major international deployment of the NEES@UCLA structural monitoring equipment involved overcoming various logistical, technical, and non-technical challenges. Some of these challenges are summarized here to facilitate and improve upon future deployments, both in the US and abroad. Establishing a list of local contacts and potential collaborators prior to an earthquake is essential since local communications are less reliable and the ability to deploy rapidly is impacted. The list of local contacts should include both university researchers and professionals (engineers, architects, developers) to identify candidate structures, review building drawings and to obtain deployment permissions and building access. University students are likely to be an excellent source of local assistance for developing and maintaining deployments,
337
identifying local sources for equipment purchases, as well as finding specialty accessories needed to enable connection to local power/internet. For each location, establishing a list of research topics that can be addressed by collecting aftershock data. This list provides the justification for the monitoring and will speed up the process of obtaining funding to support the effort. For foreign deployments, import permits for US owned equipment can be facilitated if letters from local collaborators, in both English and the local language, can be shown upon entering the country. To avoid customs and VAT, letters should officially state that the imported equipment is not for sale, will only be in the country temporarily, and will be taken out of the country following the collaborative research effort. Transporting equipment via airline (i.e., as personal luggage) is significantly cheaper (up to 5) than shipping the same equipment with a professional carrier (e.g., FedEx, UPS). Take advantage of personnel with high-level frequent flyer status (more pieces of luggage and higher weight allowances). Label every piece of luggage with its contents and maintain a copy of this list during the entire trip. Similarly, label every sensor and cable prior to departure. For long cables splitters and connectors (powered connectors if needed) may not be available locally and should be packed and shipped. Shipping heavy batteries is not practical; procuring local batteries (e.g., large 12 V car batteries) as needed to power data acquisition equipment is the best option. Ensure that locally purchased batteries are charged. Use of ‘‘rapid’’ charging might help overcome this problem; however, battery damage is possible. Clearly identify participant roles and identify backup roles and expertise. In this deployment, we found it helpful to have 1 team manager, 1 technical expert (needs to be familiar with the sensors and DAQ system), 2–3 people to help install (at least one of them should be a local contributor), 2 local collaborators (faculty or practicing engineers in the respective country). Document everything you do with time-stamped (geo-stamped if possible) photographs and video.
We cannot overstate the value of local collaborators. In this deployment, our local contacts took responsibility selecting buildings, providing access to drawings and other technical information, and conducting negotiations with building owners and managers.
6. Summary and conclusions Over a period of several weeks, four reinforced concrete buildings located in Santiago, Chile were instrumented with accelerometers and LVDTs to record aftershock response data. Two out of the four buildings were selected to conduct simple system analyses as well as some preliminary modeling studies to show what building parameters/behaviour can be easily retrieved within a short amount of time following the seismic event. Both buildings were 10 story structures with different lateral force resisting systems, i.e. a moment frame system and a shear wall system. Accelerations in both buildings increased by a factor of 2.5–4 in EW or NS directions between ground and roof floors. First mode building frequencies were about 0.9–1.2 Hz with corresponding building periods of T = 0.83–1.1 s. A strong sensitivity to torsional accelerations was observed in the damaged shear wall building. An instrumented reinforced concrete shear wall showed that shear displacements were stronger by a factor of five compared to measure flexural deformations during the aftershock. General shear cracking occurred predominantly in wall areas when openings in floor levels
338
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
above or below the respective floor were discontinued. Simple modeling studies of the 10 story residential shear wall building using the computer program ETABS provided comparisons in building behaviour when implementing stiffness reduction factors according to FEMA 356. The model showed that the natural period was best captured under the assumption that elements with limited cracking were assigned 80% of the stiffness, while elements with severe damage or failure were assumed to have no contribution to the overall structural stiffness of the building. All data are stored at the NEES project warehouse and form a first initiative to create a public database for building response data to seismic events. Valuable practical lessons were learned during this effort and are shared with the reader. Acknowledgements This instrumentation study was funded through an award granted by the U.S.-National Science Foundation (NSF RAPID – CMMI 1040574). Supplementary travel support was provided by the U.S.-Earthquake Engineering Research Institute (EERI) as part of the 2010 Chile Earthquake reconnaissance efforts. The authors would like to acknowledge the Network for Earthquake Engineering Simulation (NEES) which provided the instrumentation and data acquisition equipment for the field study as part of the NEES-SHARED USE program. We would also like to acknowledge the help of a team consisting of local and international professionals and students without whom this process would not have been made possible at the speed and organizational management at which it was completed: Dr. Robert Nigbor (NEES@UCLA), Dr. Alberto Salamanca (NEES@UCLA), Matias Chacon, Javier Encina and Joao Marques (currently/formerly at Pontificia Universidad Cathólica de Chile), Dr. Marc Sereci (Digitexx) and Aditya Jain (ON Semiconductor), Aziz Akhtary (formerly at CSUF) and Pedro Quezada (formerly at University of Los Andes, Santiago). References [1] Los Angeles Tall Building Seismic Design Council – LATBSDC. An alternative procedure for seismic analysis and design of tall buildings located in the los angeles region.; 2011. [2] Los Angeles Amendment Building Code. ; 2011. [3] California Strong Motion Instrumentation Program (CSMIP). California geological survey, california department of conservation.
[8] Bradford SC, Clinton JF, Favela J, Heaton TH. Results of millikan library forced vibration testing, technical report: Caltech EERL-2004-03. California Institute of Technology; 2004. [9] Doebling SW, Farrar CR, Prime MB, Shevitz DW. Damage identification and health monitoring of structural and mechanical systems from changes in their vibration characteristics: A literature review. Los Alamos National Laboratory report LA-13070-MS; 1996. [10] Celebi M, Sanli A, Sinclair M, Gallant S, Radulescu D. Real-time seismic monitoring needs of a building owner – and the solution: a cooperative effort. Earthquake Spectra 2004;20:333–46. [11] Naeim F. Performance of 20 extensively-instrumented buildings during the 1994 Northridge earthquake. Struct Des Tall Build 1998;7(3):174–94. [12] Goel RK, Chopra AK. Period formulas for moment resisting frame buildings. J Struct Eng, ASCE 1997;123(11):1454–61. [13] Goel RK, Chopra AK. Period formulas for concrete shear wall buildings. J Struct Eng, ASCE 1998;124(4):426–33. [14] Earthquake Engineering Research Institute (EERI). The Mw 8.8 ChileEarthquake of February 27, 2010. EERI Special Earthquake Report. June 2010. (May 2010). [15] American Concrete Institute (ACI) 318. Building Code Requirements for Reinforced Concrete, American Concrete Institute, Detroit, Michigan; 1995. [16] NCh433.Of96. Earthquake resistant design of buildings (in Spanish), Instituto Nacional de Normalización, INN 1996; 1996. p. 43. [17] Massone LM. Fundamental principles of the reinforced concrete design code changes in chile following the Mw 8.8 earthquake in 2010. Eng Struct 2013;56:1335–45. [18] Massone LM, Bonelli P, Lagos R, Lüders C, Moehle J, Wallace JW. Seismic design and construction practices for RC structural wall buildings. Earthquake Spectra 2012;28(S1):S245–56. [19] Experiment-1: Anne Lemnitzer, Leonardo Massone, Derek Skolnik, Alberto Salamanca, Juan Carlos De La Llera, John Wallace. Ambient Vibration Testing of 10 Story Office Building (Building 2). Network for Earthquake Engineering Simulation (NEES) (distributor). Dataset; 2013. doi: http://dx.doi.org/10.4231/ D3PK0722Z. [20] Experiment-2: Anne Lemnitzer, Leonardo Massone, Derek Skolnik, Alberto Salamanca, Juan Carlos De La Llera, John Wallace. Ambient Vibration Testing of 25 Story Office Building (Building 1). Network for Earthquake Engineering Simulation (NEES) (distributor). Dataset; 2013. doi: . [21] Experiment-3: Anne Lemnitzer, Leonardo Massone, Derek Skolnik, Alberto Salamanca, Juan Carlos De La Llera, John Wallace. Ambient Vibration Testing of 10 Story Residential Building (Building 3). Network for Earthquake Engineering Simulation (NEES) (distributor). Dataset; 2013. doi: http:// dx.doi.org/10.4231/D3JS9H78N. [22] Experiment-4: Anne Lemnitzer, Leonardo Massone, Derek Skolnik, Alberto Salamanca, Juan Carlos De La Llera, John Wallace. Ambient Vibration Testing of a 5 Story Office Building (Building 4). Network for Earthquake Engineering Simulation (NEES) (distributor). Dataset; 2013. doi: http://dx.doi.org/10.4231/ D3F18SF4T. [23] International Building Code. International Code Council, Whittier, CA; 2010. [24] Wallace J, Massone L, Bonelli P, Dragovich J, Lagos R, Lueders C, Moehle J. Damage and implications for seismic design of RC structural wall buildings. Earthquake Spectra 2012;28(S1):S281–99. [25] MATLAB. MATLAB Version 7.10. Natick, Massachusetts: The Mathworks Inc.; 2010. [26] Hejal R, Chopra AK. Earthquake analysis of a class of torsionally-coupled buildings. Earthquake Eng Struct Dyn 1989;18:305–23. [27] Moroni MO, Sarrazin M, Soto P. Behavior of instrumented base-isolated structures during the 27 February 2010 Chile Earthquake. Earthquake Spectra 2012;28(S1):S407–24. [28] ETABS. ETABS Version 9.1.4. Computers and Structures, INC. Walnut Creek, California; 2013. [29] FEMA 356. Prestandard and Commentary for the Seismic Rehabilitation of Buildings, prepared by the SEAOC, ATC, and CUREE Joint Venture for the Federal Emergency Management Agency, Washington, D.C. ; 2000.
Contents lists available at ScienceDirect
Engineering Structures journal homepage: www.elsevier.com/locate/engstruct
Aftershock response of RC buildings in Santiago, Chile, succeeding the magnitude 8.8 Maule earthquake Anne Lemnitzer a,⇑, Leonardo M. Massone b, Derek A. Skolnik c, Juan C. de la Llera Martin d, John W. Wallace e a
University of CA, Irvine, Dept. of Civil & Environmental Engineering, 4149 Eng. Gateway, Irvine, CA 92697, USA University of Chile, Santiago, Blanco Encalada 2002, Santiago, Chile Kinemetrics Inc., 222 Vista Avenue, Pasadena 91107, CA, USA d Department of Structural and Geotechnical Engineering, Pontificia Universidad Catolica de Chile, and National Research Center for Integrated Natural Disaster Management CONICYT/FONDAP/15110017, Vicuna Mackenna 4860, Santiago, Chile e University of California Los Angeles, 405 Hilgard Ave, Los Angeles 90024, USA b c
a r t i c l e
i n f o
Article history: Received 10 November 2013 Revised 1 July 2014 Accepted 2 July 2014
Keywords: Seismic monitoring Reinforced concrete structures Instrumentation
a b s t r a c t Between March 13th and 28th 2010, a team of U.S. researchers, professionals, and local collaborators instrumented four reinforced concrete (RC) buildings in Santiago, Chile, to measure aftershock response data following the February 27, 2010 MW 8.8 earthquake. The selected buildings, designed according to NCh433.Of96 (similar to ACI 318-95), represent typical construction, e.g. moderate to high rise office buildings with large office space and inner core shear walls and mid-rise residential shear wall buildings. Two of the instrumented buildings were undamaged, whereas one building suffered only minor nonstructural damage and the fourth building exhibited more significant structural damage such as column buckling and shear wall cracking. Instrumentation consisted mainly of uni-axial and tri-axial accelerometers as well as some displacement transducers. Records for several aftershocks were captured during a period of one month. Collected data were processed and system identification algorithms were used to determine dynamic building- and modeling parameters (e.g. building periods, acceleration amplification, inter-story drifts, building rocking and torsion, and assessment of diaphragm in-plane rigidity). A comparison between linear elastic models used in engineering practice and measured data was performed for one building. Information related to the instrumented buildings, data collected, and analytical results is presented, along with practical lessons learned conducting monitoring studies in the aftermath of this very strong earthquake. Ó 2014 Elsevier Ltd. All rights reserved.
1. Introduction Building instrumentation has become essential in seismically active areas to gain insight into building behaviour under earthquake loading. For example, the city of Los Angeles has included modest levels of instrumentation in the applicable building code since 1965. Since 2010, more extensive levels of instrumentation are required for tall buildings designed using non-linear response history procedures of ASCE 7 Chapter 16 [1,2]. The California Strong-Motion Instrumentation Program [3] and the Advanced National Seismic System [4] are tasked with operating larger instrumentation programs for the state of California and the Federal level respectively. A detailed history of building
⇑ Corresponding author. Tel.: +1 (310)986 5255. E-mail address: [email protected] (A. Lemnitzer). http://dx.doi.org/10.1016/j.engstruct.2014.07.003 0141-0296/Ó 2014 Elsevier Ltd. All rights reserved.
instrumentation programs within the United States is available in Celebi [5]. Data from permanently and temporarily instrumented buildings have enabled important studies in system identification and model updating [6,7] damage detection and long-term health monitoring [8–11] and have led to important code provisions such as fundamental period formulas [12,13]. The goal of these programs and provisions is to monitor various building construction types to gain better understanding of building performance under dynamic loading and enable the improvement of current design codes and practices. Damage and failures observed after the MW 6.7 Northridge earthquake in 1994 led to significant changes to buildings codes; however, since 1994, the United States has not experienced a large magnitude earthquake (>7.0); therefore, major changes over the past decade or so have relied on observations from, and studies of, earthquakes occurring outside of the US. While permanent instrumentation programs (such as those
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
previously described) are gaining traction around the world, their evolution is slow due to the high costs associated with equipment installation and maintenance, as well as the lack of building instrumentation in locations that have experienced strong shaking. In turn, rapidly deployable temporary instrumentation programs in areas with frequent seismic events provide a unique opportunity to collect significant amounts of data over a short period of time. The NEES@UCLA facility includes equipment and expertise to enable rapid system deployments to enable rapid aftershock monitoring. Aftershock studies, in particular, offer excellent possibilities to capture exclusive data for buildings damaged during the main shock, such as the validation of modeling assumptions during the design process or evaluation of various retrofit schemes. Several hundred aftershocks followed the main rupture of over 550 km long on February 27th, 2010s MW 8.8 Maule earthquake in Chile. Most large magnitude shocks (M > 6.0) occurred within 24 h after the main event [14] while moderate aftershocks continued for over 5 weeks. The epicenters of the majority of aftershocks were located near the epicenter of the Mw 8.8 earthquake (i.e. near Concepcion), but were felt in radial distances of up to 300 km, including the areas of Vina del Mar/Valparaiso, Santiago and Mendoza. A team of researchers and professionals visited several reinforced concrete buildings in Chile’s capital Santiago as part of a reconnaissance effort in March 2010 and selected four buildings for temporary instrumentation over a period of 4–8 weeks. The buildings were chosen based on their availability/accessibility and design similarities to reinforced concrete structures present in the United States in order to make the measured data most beneficial for both countries. The selected structures were designed using the local, pre-MW 8.8 design recommendations for reinforced concrete buildings, which were taken from the ACI 318-95 [15] specifications in the United States, with the exception of the boundary element confinement provisions for shear walls. Chile’s design code [16] was revised between 2010 and 2011 and now includes many lessons learned from the 2010 Maule Earthquake and its reconnaissance efforts such as modifications of wall boundary transversal reinforcement, axial load limitations, damage limitations, minimum thickness requirements of confined walls. A summary of the adapted changes along with a thorough explaining of the reasoning can be found in Massone and Massone et al. [17,18]. The objective of this study was to collect structural performance data that would enable a basic analysis to quickly provide the most important building dynamic and modeling parameters such as periods, acceleration amplification factors, inter-story drifts, and rocking responses as well as torsional effects and damping ratios (if possible). It also forms the foundation to guide subsequent in depth studies such as nonlinear structural modeling and comprehensive system identification analyses. This study also forms an initiative to create a unique database of structural input and response data from aftershocks and main shocks. It represents one of the first efforts to organize a spontaneous team to visit a foreign country immediately following a seismic event that imports of all
325
instrumentation and data acquisition technologies from the United States. Valuable practical lessons were learned during this effort and applications and limitations with monitoring studies in foreign countries are shared with the reader and are used to enhance the success of future similar efforts. All data of this instrumentation effort are publicly available at the NEES project warehouse [19–22] which can be found at https://nees.org/warehouse/ report/project/941. 2. Selected case studies The four buildings selected for instrumentation are reinforced concrete structures located in downtown Santiago and are depicted in Fig. 1. Buildings 1 and 2 are situated in Santiago’s business district ‘‘Las Condes’’ in the North-East of the city. Building 3 is located in a prominent residential district ‘‘Vitacura’’ and Building 4 was located in Santiago’s Huechuraba district, an area known for very soft soil conditions. The following paragraphs will provide a brief overview of the building’s structural systems along with observations of successful performance or damage during the main shock of the February 27th, 2010 earthquake. Particular focus is paid to Buildings 2 and 4, since most measurements and analyses were conducted for those structures. Buildings 1 and 4, are explained in brief; the reader will be referred to the NEES project warehouse for further information and ambient vibration data. Building 1 (Fig. 1) is a 25 story high-rise office building, with additional five subterranean parking levels, whose lateral force resisting system consists of two inner-core shear wall elevator shafts complemented by an external beam-column frame system. No earthquake damage (neither structural nor architectural) was observed in this building (Fig. 2). Building 1 was instrumented with four tri-axial accelerometers on the four exterior corners of the ground floor and four corresponding accelerometers on the roof (25th floor). Instrumentation was installed over a period of four days, measuring four aftershocks with rather small peak accelerations between 0.6 cm/s2 and 2.25 cm/s2 (0.057–0.23%g). Building 1 was the first building instrumented and challenged the instrumentation team with many issues related to the equipment installation such as the continuation and interference of business operations, scouting most suitable locations for sensors and cable runs (logistically and structurally), running cables over the entire building height of (84.15 m/25 stories), and introduction of signal noise stemming from long cables, logistical aspects and equipment purchase on site (battery powered equipment and overnight recharging, outlet power converters). Several hours of ambient vibration, as well as two small aftershocks, were captured in Building 1 and are stored on the NEES project warehouse website along with photographs and instrumentation plans. Building 2 (shown in Fig. 1) is a 10 story reinforced concrete building with six levels of subterranean parking. The building was under interior construction at the time of instrumentation and provided several unoccupied floors for easy access and installation of
Fig. 1. Instrumented Building 1–4 (left to right) in Santiago, Chile.
326
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
Fig. 2. Building 1, no damage detected on façade or soft story.
sensors. The structural system consisted of small inner core shear walls formed by the elevator shaft and staircase (thickness ranging from 20 to 30 cm) and an external beam-column framing system with 0.82 0.4 m columns in the long direction and 0.5 m 0.4 m columns in the short direction spaced at 2.5 m and 3.6 m center to center in each direction respectively. Column dimensions on the 10th floor reduced to 0.5 m 0.4 m in the long direction. Typical column reinforcement consisted of A630-420H steel (fy = 420 MPa), #8–10 bars around the column periphery and transverse reinforcement of #3–4 hoops spaced at d/2. The total floor area of Building 2 is approximately 3209 m2. The office
building is a representation of a structure in the Occupancy Category 2 [23]. Similar to Building 1; the structure of this building did not suffer any structural damage during the Mw 8.8 earthquake or its subsequent aftershocks and represents a success story for well-designed RC buildings. Fig. 3a and b shows the location of Building 2 in the business district of Las Condes, as well as an interior view of the unoccupied office space on floor 10. Building 3 (Fig. 1) was located in a newly developed luxurious residential district of Santiago in the north-east part of the city and represents a 10-story reinforced concrete shear wall building with one level of underground parking. Fig. 4(a)–(c) shows the
Fig. 3. (a and b) Location of Building 2 (left) and view of interior office space used for instrumentation.
Fig. 4. (a–c) Building 3 front view (a), side view (b) and damage at 1 story level (c).
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
327
Fig. 5. First floor plan view with more severely damaged elements.
building in front and side view. Fig. 5 presents a structural plan of its first floor. Floors 3–8 were typical floors and maintained many of the configurations shown in the first floor. At the time of deployment (March 2010) only floors 2 and 9 were unoccupied and chosen for instrumentation. Aside from floors 2 and 9, additional sensors were placed at the subterranean level and the roof. A total of 10 accelerometers and four linear voltage differential transducers (LVDTs) were installed in the building. The building suffered moderate structural damage at the subterranean parking level which consisted mainly of spalling of the concrete cover, section cracking and longitudinal rebar buckling at the column and wall to ceiling interface (see Fig. 6). Damage on the ground level included column failure in two structural columns supporting the second floor (Figs. 5 and 7) which
was introduced through large axial compression forces. Similar instabilities were observed by Wallace et al. [24] for other RC structures in Chile. Significant shear cracking was noticed on interior and exterior shear walls (Fig. 8) Smaller shear cracks were detected in floors 2–10 and reoccurred consistently in exterior walls when window (or door) openings were discontinued in floors located above or below (i.e., producing an irregular opening pattern over the building height). The structure is a commonly used type for residential buildings in Chile. Thicknesses of shear walls and slabs were generally 17 cm and 15 cm, respectively. Typical material properties are a concrete strength of H30 (f0 c = 25 MPa) and reinforcing steel with Grade A630-420H (equivalent to US Grade 60, fy = 420 MPa). The building was designed according to NCh433.Of96, following most of the ACI 318-95 requirements
Fig. 6. (a and b) Moderate damage at underground parking level Building 3.
328
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
were measured but no aftershocks were captured in this building during the instrumented period. This data, along with structural plans, is available on the project warehouse for future studies and may be particularly beneficial during investigations in future seismic events, in particular for soil structure interaction studies, given the soft soils at the site of Building 4. The remainder of this manuscript will limit its focus on Buildings 2 and 3 for which the most aftershocks with quality data were obtained. Both buildings incorporate typical lateral load resisting systems (moment frames or shear walls) and provide a good representation of the overall data collection effort and building analyses. Although measured accelerations were small, recordings were sufficient to conduct fundamental analyses of the respective buildings and provide a comprehensive insight into important building parameters. 3. Instrumentation
Fig. 7. Two column failures at ground level due to axial and shear loading.
with the exception of the boundary element confinement of shear walls. As it can be seen in Fig. 5, the area of walls was distributed almost equally in both orthogonal directions of the building. The wall to floor area ratio in each direction was about 1.5%, which is typical but could be considered in the lower end of wall to plan ratios in Chile [18]. The static loads in the structure at the time of the measurements correspond to the self-weight (dead load) including structural and non-structural elements, and certain live loads which were expected to be relatively small since the buildings was partially uninhabited during monitoring. Building 4 (Fig. 1) was a five story moment frame office building which was under interior construction (partitions, ceilings) at the time of instrumentation. Large amounts of ambient vibrations
The instrumentation in Buildings 2 and 3 consisted of uniaxial and triaxial Episensor accelerometers with a measurement range of ±4 g. Additionally, Building 3 was instrumented with several linear voltage differential transducers (LVDTs) installed at the damaged shear wall (see Fig. 9). All data were recorded with Quanterra Q330 digitizers and processed with MATLAB [25]. Data were recorded at a sampling rate of 200 Hz, and recording was triggered when accelerometer readings exceeded a threshold of 0.005 g. Each recording included 15 s of pre- and post-event data. Several hours of ambient vibration data were also recorded during day and night time to establish baseline behaviour. The monitoring system was powered using large 12 V batteries which were locally purchased and charged onsite. All monitoring equipment belonged to NEES@UCLA and was brought into Chile using a temporary instrumentation permit issued by the customs and immigration office upon arrival in Santiago. Fig. 10 shows a 3D instrumentation layout of Building 2 with the sensors indicated on floors 1 and 10. Fig. 11 shows the plan layout and exact location of the accelerometers installed on the respective floors. Floor one was instrumented with three uniaxial accelerometers in each corner of the building and floor 10 was instrumented with one triaxial accelerometer in each corner. All instrumentation was directly mounted on the reinforced concrete slabs. Four aftershocks with moment magnitudes ranging between 4.7 and 5.1 were measured in Building 2. The amplitude of translational acceleration over the multiple aftershocks ranged from 0.001 to 0.01 g. Fig. 12 shows the simplified 3D schematic of Building 3. This structure was instrumented with one triaxial accelerometer at the center of the building on the basement slab, i.e. on the concrete slab at underground parking level, three triaxial accelerometers on the second floor (installed on tile floors, or directly on concrete slab), one triaxial accelerometer at the ninth floor (on concrete
Fig. 8. (a and b) Shear wall damage occurring at discontinuity of window openings on the 1st level, outside (a) and inside view (b).
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
329
Fig. 9. Instrumentation of damaged shear wall in Building 3.
Fig. 12. Instrumentation layout in Building 3.
slab) and three uniaxial accelerometers at the roof (installed at the concrete slab). A layout of the floor plan with the instrumentation is illustrated in Fig. 13. Between March 18th and May 5th, a total of 30 aftershocks were recorded in Building 3 with peak ground horizontal accelerations ranging from 0.5 cm/s2 to 7 cm/s2 (0.05–0.7%g). A total of 22 aftershocks were carefully checked for the proceeding data evaluation and analysis. 4. Results 4.1. Aftershock recordings in Building 2
Fig. 10. Instrumentation layout in Building 2.
A M4.7 aftershock recorded on March 25, 2010 at 10:59:35 am local time (12:59:35 pm UTC time) was selected to study the seismic response of Building 2 (Fig. 14). The epicenter was located about 17 km north of Pichilemu, which is located about 200 km south of Santiago, at a depth of 34.5 km. This epicenter coincides with the region of largest energy liberation during the earthquake. Prior to processing the measured structural accelerations, the data set was adjusted via filtering with linear bandpass filters and baseline corrections to eliminate noise. Several different sets of measured accelerations at the corners were used to calculate the three dimensional (two translational
Fig. 11. Location of instrumentation in Building 2.
330
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
Fig. 13. Location of instrumentation in Building 3.
and torsional) slab motions at a central point assuming that each slab behaves as a rigid diaphragm. Results of this analysis showed that the rigid diaphragm assumption is sound. Fig. 15 describes the relationship between local translational acceleration at any given point on a floor (ui and vi corresponding to the ith sensor) and the central (geometric) story motions (u0, v0, and h0). First floor accelerations with maximum values of 1.2 cm/s2 both, in EW and NS directions were found. Floor 10 shows center accelerations of 3.5 cm/s2 and corner accelerations of 4.5 cm/s2 in the EW direction, and center accelerations of 2.8 cm/s2 and corner accelerations of 5.0 cm/s2 in the NS direction. The amplification of acceleration between floor one and ten is in the range of 2.0–4.5 (center and corner respectively). As expected through comparison of floor corner and center accelerations, torsion was apparent on the 10th floor. Fig. 16 illustrates the maximum torsional acceleration on the first and tenth floor of 1.273 104 rad/s2 and 3.386 103 rad/s2 respectively. Double numerical integration of recorded accelerations enabled the calculation of story displacements. In the E-W direction, maximum displacements of 0.017 cm and 0.104 cm were measured for floors one and ten, respectively. The total inter-story drift between floors one and ten was 0.088 cm. In the N-S direction, maximum displacements of 0.015 cm and 0.061 cm were recorded for floor one and ten respectively, resulting in an inter-story drift of 0.046 cm, which is approximately half of the drift measured in the orthogonal direction. Displacement measurements confirm minimal torsion at level one, but twisting is observed on floor 10. The maximum relative torsional displacement on the tenth floor is 4.50 105 rad. The XY displacement of the center of floor ten is shown as particle motion – time history for the duration of the aftershock in
Fig. 17. The total displacement range in the EW direction spans over 1.8 mm and the total displacement range in the NS direction is approximately 1.35 mm. Fast Fourier Transforms (FFT) (Fig. 18) were computed to estimate buildings transfer functions (Fig. 19). Transfer functions were developed for E-W, N-S and torsional direction to identify the dynamic parameters of the first and second modes of the structure. From Fig. 19, it can be seen that the first natural mode at about 0.89–0.91 Hz (T = 1.1 s) is predominantly NS but evident in both EW and torsional directions, thus it is most likely a coupled NS-torsional mode with a close second mode in the EW directions appearing at 0.96 Hz (T = 1.04 s). The identification of coupled modes is important because it means that the lateral forces developed in the structure’s resisting elements are generally higher than they would be for a typical symmetrical building with planar-only, well-spaced modal responses [26]. Such findings should be taken into consideration during construction of analytical models and interpretation of results. A possible third mode of the structure is evident in the NS direction at 2.35 Hz (T = 0.43 s). Fig. 20 shows a simple analysis using aftershock and ambient data to investigate the relationship between the frequency response of the building in EW and NS direction and the intensity of the ground movement in form of peak floor accelerations at the 10th floor of Building 2. A clear trend of building frequency reduction with increasing floor acceleration can be observed, which means that as accelerations increase, the overall building period becomes larger. This apparent nonlinearity has been affirmed in countless similar studies and is typically explained by citing the nonlinear behaviour of soil–structure interaction and the reduced contribution of non-structural components [6].
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
331
Fig. 16. Acceleration time histories for Building 2.
Fig. 14. Map of approximate aftershock locations.
Fig. 17. Relative displacement–time history for 10th floor particle motion at the center of the slab – Building 2.
Fig. 15. Terminology of displacements and rotations in @ Center & Corners.
4.2. Aftershock recordings in Building 3 A 5.8 Magnitude aftershock which occurred on May 2nd, 2010 at10:52:30 local time (14:52:39 UTC) at a depth of 32.9 km, 12 km north of Pichilemu (see Fig. 14, http://www.sismologia.cl) was identified as the strongest aftershock in terms of absolute accelerations, and selected for detailed processing in Building 3. The peak ground accelerations (PGA) reached about 4.3 cm/s2 in EW and 7.2 cm/s2 in NS direction. The structural accelerations measured in the building are presented in Fig. 21 and Table 1.
Again, a band-pass filter with pass frequencies ranging from 0.1 to 25 Hz was implemented for data pre-processing. The amplification of accelerations from ground to roof was calculated to be 5.3 in the EW direction and 3.4 in the NS direction. Fig. 22 shows the amplification of the roof peak acceleration vs. PSA for more than 20 aftershocks recorded in Building 3. There is a clear linear trend of amplification increase up approximately 5 cm/s2, after which it tends to remain constant or even decrease. Similar results were observed by Moroni et al. [27] for a permanently instrumented, four story, RC/masonry building in the Andalucia Community in Chile. Measurements during the main shock of the 2010 Maule earthquake indicated an increase in roof amplitude up to peak ground accelerations of 20 cm/s2 which a constant ratio thereafter [27].
332
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
Fig. 18. Fast Fourier Transforms for Building 2 (FFT).
Fig. 20. Building 2 – EW and NS frequencies vs. peak floor accelerations (PFA). Fig. 19. Building 2 transfer functions.
The exact location of all sensors in Building 3 was previously presented in Fig. 13. Given the larger amount of sensors installed, this building allowed for more analyses than the previous case. Selected parametric studies and simply modeling analyses are shown hereafter. Second floor accelerations in Building 3 were estimated using a least square analysis since the three triaxial sensors installed at that level (see Fig. 13), over-determined the assumed 3 translational DOF of the floor (x, y and z). Given the instrumentation configuration on the 2nd floor, torsional and translational accelerations were determined. The maximum torsional acceleration reaches about 0.004 rad/s2. The largest translational acceleration measurement in floor 2 reached about 10 cm/s2 (1%g) in both NS and EW direction. Thus, a sensor located 10 m away from the central sensor would result in an increment of translational acceleration of 4 cm/s2 (0.004 * 1000) due to torsion, which corresponds to about 40% of the translation of the central sensor, indicating that the torsional component has a significant impact in the response in Building 3.
Fig. 21 (b) shows the EW and NS absolute displacements components of all instrumented floors. The displacements were determined by double-integrating the accelerations of each level and applying a high-pass filter with 0.2 Hz low end corner frequency to overcome residual displacements. The peak displacement increased from about 0.08 cm in the NS direction and about 0.04 cm in the EW direction on the ground floor to 0.23 cm in NS and 0.21 cm in EW direction at the roof level, indicating a displacement amplification over the entire building height of 4. Displacements amplitudes at the 9th floor and the roof were found to be very similar in magnitude (Table 1). Fig. 23 shows particle motion as absolute displacements in the NS vs. EW direction measured for various levels in the building. It can be seen that floors ‘Ground’, 2 and 9 present patterns that move over the entire plane (uniformly circular). The roof level in turn shows a clear tendency to move along a diagonal path, which is consistent with the previous observation of the roof’s EW and NS components. The roof sensor, which is eccentric relative to the other sensors, is influenced by a torsional mode given the synchrony between both displacement components. Fig. 24 presents
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
333
Fig. 21. (a and b) Story accelerations (left) and displacements (right) in EW and NS direction for Building 3.
Table 1 Building 3 – floor accelerations and displacements.
Ground 2nd 9th Roof
Accelerations (cm/s2)
Displacements (cm)
EW
NS
EW
NS
4.3 8.5 8.93 22.6
7.2 13.3 14.5 24
0.05 0.07 0.18 0.21
0.08 0.11 0.2 0.23
Fig. 23. Particle motions Building 3.
Fig. 22. Roof acceleration amplification vs. PSA in EW and NS direction in Building 3.
the transfer functions of the relative displacement response vs. the normalized ground acceleration determined at each instrumented floor. In order to smooth the observed response a Welch procedure was implemented. From the analysis, the main frequency (f) and damping ratio (b) were determined to be f = 1.09 Hz and b = 3.0% in NS direction, and f = 1.2 Hz, b = 2.2% in the EW direction. The largest amplitude of the relative displacement response occurred at a frequency of 1.09 Hz in the NS direction at the roof level. The maximum absolute accelerations of selected floors were compared with pseudo spectral accelerations (PSA) determined for a single degree of freedom system (SDOF) using the measured building parameters (natural frequency and damping ratio) that were observed in the corresponding direction (i.e., NS: f = 1.09 Hz, b = 3.0%; EW: f = 1.2 Hz, b = 2.2%). The results for both directions are presented in Fig. 25. A trend line (best fit analysis) is included to show the linear trend between the maximum floor
accelerations and the intensity of the 22 aftershocks used for this analysis. Although, increases of PSA result in increases of maximum acceleration at different levels, the 2nd floor presents little correlation with PSA (R2 < 0.4). Fig. 26 shows a comparison of the maximum relative displacements of various floor levels with the results of the pseudo spectral analysis. Good correlation is observed between the displacement and PSA (R2 > 0.8) for all levels of Building 3. Additional studies were conducted for absolute displacements and showed consistent results when compared against peak ground displacements (PGD). It can also be seen that close trends are obtained for the 9th floor and roof. All curves aim for zero displacement at zero PSA. A comparison of the overall natural building frequency (3 modes) in NS direction vs. pseudo spectral accelerations is shown in Fig. 27 using different aftershocks and ambient vibration data. Pseudo spectral accelerations were determined using first mode natural periods and damping as described above. For ambient vibration cases, PSA values were assumed zero. Although there is scatter, the trend line in Fig. 27 shows a clear tendency of frequency reduction with increasing magnitude of the ground movement, which is consistent with observations made for Building 2 in Fig. 20. This trend may be associated with the stiffness variation due to opening and closing of small cracks in the structural elements of the building.
334
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
Fig. 24. Transfer functions Building 3.
Fig. 26. Comparison of relative floor displacements with pseudo spectral accelerations (PSA). Fig. 25. Comparison of measured floor accelerations with pseudo spectral accelerations (PSA).
The instrumentation in Building 3 included the installation of linear variable differential transducers (LVDTs) across a damaged shear wall on the first floor of the building. A photograph of the
wall and the sensor instrumentation was previously shown in Fig. 9. One pair of LVDTs was installed in diagonal configuration to enable the calculation of shear deformations during the aftershock. Another pair of vertical LVDTs was placed at both ends of the wall, spanning over the entire story height to determine the
335
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
Fig. 27. Building frequency vs. pseudo spectral accelerations (PSA).
rotation due to flexural deformations. Since deformations are small, the diagonal sensors allowed for estimating the shear component of the first story lateral deformations in the wall plane. The flexural component was estimated by assuming a concentration of such deformation at wall base and that overall deformations in the wall are dominated by shear. A time history of shear and flexure displacements during the March 25 aftershock is presented in Fig. 28. The results indicate that the deformations are dominated by the shear component reaching about 0.2 mm of lateral deformation, whereas the flexural component reaches about 0.04 mm, which is only 20% of the shear component. 4.3. Simple modeling studies performed in Building 3 The amount of building information and data measurements available for Building 3 enabled a basic modeling study using the structural program ETABS [28]. The building was modeled as a 3D structure based on the original construction documents. All elements were considered linear elastic, and diaphragms were considered rigid in their plane. In order to compute the stiffness of the elements, gross concrete sections were used for the analysis, but later modified to account for degradation of stiffness in order to represent the elements in the state in which the aftershock measurements took place (i.e. damaged state). The structure was assumed to be fully fixed in all degrees of freedom at the basement–ground interface.
The gravity loads were uniformly distributed across all floor levels. A superimposed dead load of 0.1 ton/m2 (excludes the self-weight of the structure) and a live load of 0.2 ton/m2 were calculated and applied in the analytical model. The total gravity load per floor acting in the structure (dead and live, including self-weight) results in a distributed weight of approximately 1.0 ton/m2. The process of construction, exposure, maintenance, ageing and loading (in particular seismic loading), contributes to the variation of material properties in a building; for this specific case: the properties of the reinforced concrete. Considering the limitations of the available information, recommendations by FEMA 356 [29] were used to incorporate reduction factors (RF) in the ETABS model. Hereby, factors that consider different levels of cracking for the respective elements were most crucial. The reduction factors are shown in Table 2. For the purposes of this analysis, all columns stiffness’ values were reduced by 50% (meaning I = 0.5Ig, where Ig, is the moment of inertia of the gross section). Walls stiffness reduction was varied and a sensitivity study for this parameter was conducted. To enable comparisons, a baseline model (Model 1 – B) was established using the standard values in Table 2. The baseline model included a live load of 0.2 ton/m2 and a dead load of 0.1 ton/m2 in addition to its regular self-weight. For the structural stiffness of the columns a reduced value of 0.5Ig was used, and for the structural stiffness of the walls a reduced value of 0.8Ig was used. The stiffness of the severely damaged elements on the first floor was set to zero. (i.e., elements highlighted in Fig. 5). In a second model (Model 2 – IL), the stiffness parameters were kept identical, but gravity loading was increased by 0.2 ton/m2 (live loads only) which resulted in a total increase of the mass of the structure by about 20%. A third model was developed in which the loading assumptions corresponded to the baseline model, but the wall stiffness was varied (Model 3 – WS). In this model, all other stiffness reductions were removed (RFs = 1.0), all vertical elements were treated as walls and a sensitivity study of the wall stiffness alone was investigated. The severely damaged shear wall and the columns shown in Fig. 5 at the first floor were considered a zero stiffness elements. All other walls were assigned reductions of 1.0Ig (called WS100), 0.8Ig (WS80), 0.6Ig (WS60) or 0.4Ig (WS40). The ETABS model was run for all three cases and subcases and modeling results were compared with measured and back-calculated structural parameters from the monitoring effort. Table 3 shows a summary of the fundamental period of Building 3 determined for the various cases via ETABS, which corresponds to oscillations in the NS direction. The experimental fundamental period determined from building instrumentation revealed a value of 0.9 s. Analytical results show that the baseline model (Model 1 – B) is in closest agreement with the measure results. The model with the modestly increased gravity loading (Model 2 – IL) also increased the period of the structure, but maintains the same level of error as the baseline model. The sensitivity study (third case studies Models 3 – WS) indicates that the error increases significantly (up to 20% compared to measurements) when no reduction in stiffness is considered (WS100). When only the wall stiffness is Table 2 Reduction factors (based on Tables 6 and 5 – FEMA 356).
Fig. 28. Shear and flexural components of total wall displacements.
Element
Reduction factors
Beams Post-tension beams Columns – compression > 0,5 AgFc0 Columns – compression < 0,3 AgFc0 Walls Crack walls Slabs
0.5 1 0.7 0.5 0.8 0.5 0.5
336
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
Table 3 Fundamental periods of analytical model cases. Model Model Model Model Model Model Model
Fundamental period (s) 1 2 3 3 3 3
– – – – – –
B IL WS100 WS80 WS60 WS40
0.89 0.93 0.72 0.88 0.95 1.07
reduced to the original FEMA value, which is 80% (WS80), the period changes to 0.88, indicating that the modification of the flexibility of other elements different than the walls is small. This result is expected given that the structure has only very few columns and beams and is modeled like a rigid diaphragm in Etabs. A further stiffness reduction to 60% showed a continued increase in the fundamental period (T for WS60 = 0.95 s). The model with stiffness reduction to 40% showed a 19% error compared to the measured results and the model period is approximately 1.07 s. Measured and analytical time histories for accelerations and displacement enabled a simple comparison of local responses within the structure. By retrieving absolute accelerations and displacements at the locations where sensors were installed in the building, data can be directly compared without further treatment and introduction of possible errors. Figs. 29 and 30 show the
absolute acceleration records for the second floor, the ninth floor and the roof of the building. For the ninth and roof level only the NS direction is shown. In general, the model results provide a good resemblance of the shape and magnitude of sensor measurements. This indicates that the fundamental period as well as local response of the structure is adequately captured. Floor two (Fig. 29) shows a nice agreement in shape and length of the acceleration record, even though amplitudes provided by the model appear slightly less than recorded data (predominantly in the NS direction). Fig. 30 presents the comparison of data and model results for the 9th floor and similar observations can be made. The measured results tend to show a slightly smoother time history while the model shows a more distinct phase-like acceleration variation. A data-analysis comparison of the roof level showed that the amplification of the acceleration at higher floor levels is also captured in the model with increasing absolute peak accelerations from about 7 cm/s2 at the 2nd floor to about 20 cm/s2 at the roof. Lateral deformation measurements recorded at the damaged wall can be compared to the predictions from the model (baseline model was used in this case). The measurements indicate that most of the deformation originates from shear, which is consistent with the failure mode. Fig. 31 shows a comparison between model results and measurements. The top figure depicts the combined lateral deformations (shear and flexure) measured during the aftershock. The middle figure shows the modeling results, for the
Fig. 29. Absolute accelerations at the 2nd floor in E-W (left) and N-S (right) direction.
Fig. 30. Absolute accelerations at the 9th floor (left) and roof (right) in N-S direction.
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
Fig. 31. Deformation of damaged wall: top: measured total lateral deformation (shear and flexure), middle: model result for lat. deformations 0% wall stiffness (damaged wall), bottom: model results for lat. deformations – 100% wall stiffness (undamaged wall).
base model (Model 1 – B) in which the stiffness of the damaged wall was taken as zero (i.e., EI = GA = 0). The bottom figure shows the lateral deformation observed in the model when the wall was modeled with 100% of its stiffness. Results indicate that the in-situ measurements show smaller deformations than the model (Model 1 – B) predictions. This difference could be associated with the precision of the measurement equipment for such small deformation levels. A comparison between the two model results (one with 0% wall stiffness and one with 100% wall stiffness) indicated that there is a reduction in the deformation prediction, but the differences are rather modest compared to the significant change in the wall stiffness assumption. This result is expected since a rather large amount of walls were present in the structure. The displacement of the damaged wall is governed by all other connecting elements, such as other walls and slabs. 5. Lessons learned The first major international deployment of the NEES@UCLA structural monitoring equipment involved overcoming various logistical, technical, and non-technical challenges. Some of these challenges are summarized here to facilitate and improve upon future deployments, both in the US and abroad. Establishing a list of local contacts and potential collaborators prior to an earthquake is essential since local communications are less reliable and the ability to deploy rapidly is impacted. The list of local contacts should include both university researchers and professionals (engineers, architects, developers) to identify candidate structures, review building drawings and to obtain deployment permissions and building access. University students are likely to be an excellent source of local assistance for developing and maintaining deployments,
337
identifying local sources for equipment purchases, as well as finding specialty accessories needed to enable connection to local power/internet. For each location, establishing a list of research topics that can be addressed by collecting aftershock data. This list provides the justification for the monitoring and will speed up the process of obtaining funding to support the effort. For foreign deployments, import permits for US owned equipment can be facilitated if letters from local collaborators, in both English and the local language, can be shown upon entering the country. To avoid customs and VAT, letters should officially state that the imported equipment is not for sale, will only be in the country temporarily, and will be taken out of the country following the collaborative research effort. Transporting equipment via airline (i.e., as personal luggage) is significantly cheaper (up to 5) than shipping the same equipment with a professional carrier (e.g., FedEx, UPS). Take advantage of personnel with high-level frequent flyer status (more pieces of luggage and higher weight allowances). Label every piece of luggage with its contents and maintain a copy of this list during the entire trip. Similarly, label every sensor and cable prior to departure. For long cables splitters and connectors (powered connectors if needed) may not be available locally and should be packed and shipped. Shipping heavy batteries is not practical; procuring local batteries (e.g., large 12 V car batteries) as needed to power data acquisition equipment is the best option. Ensure that locally purchased batteries are charged. Use of ‘‘rapid’’ charging might help overcome this problem; however, battery damage is possible. Clearly identify participant roles and identify backup roles and expertise. In this deployment, we found it helpful to have 1 team manager, 1 technical expert (needs to be familiar with the sensors and DAQ system), 2–3 people to help install (at least one of them should be a local contributor), 2 local collaborators (faculty or practicing engineers in the respective country). Document everything you do with time-stamped (geo-stamped if possible) photographs and video.
We cannot overstate the value of local collaborators. In this deployment, our local contacts took responsibility selecting buildings, providing access to drawings and other technical information, and conducting negotiations with building owners and managers.
6. Summary and conclusions Over a period of several weeks, four reinforced concrete buildings located in Santiago, Chile were instrumented with accelerometers and LVDTs to record aftershock response data. Two out of the four buildings were selected to conduct simple system analyses as well as some preliminary modeling studies to show what building parameters/behaviour can be easily retrieved within a short amount of time following the seismic event. Both buildings were 10 story structures with different lateral force resisting systems, i.e. a moment frame system and a shear wall system. Accelerations in both buildings increased by a factor of 2.5–4 in EW or NS directions between ground and roof floors. First mode building frequencies were about 0.9–1.2 Hz with corresponding building periods of T = 0.83–1.1 s. A strong sensitivity to torsional accelerations was observed in the damaged shear wall building. An instrumented reinforced concrete shear wall showed that shear displacements were stronger by a factor of five compared to measure flexural deformations during the aftershock. General shear cracking occurred predominantly in wall areas when openings in floor levels
338
A. Lemnitzer et al. / Engineering Structures 76 (2014) 324–338
above or below the respective floor were discontinued. Simple modeling studies of the 10 story residential shear wall building using the computer program ETABS provided comparisons in building behaviour when implementing stiffness reduction factors according to FEMA 356. The model showed that the natural period was best captured under the assumption that elements with limited cracking were assigned 80% of the stiffness, while elements with severe damage or failure were assumed to have no contribution to the overall structural stiffness of the building. All data are stored at the NEES project warehouse and form a first initiative to create a public database for building response data to seismic events. Valuable practical lessons were learned during this effort and are shared with the reader. Acknowledgements This instrumentation study was funded through an award granted by the U.S.-National Science Foundation (NSF RAPID – CMMI 1040574). Supplementary travel support was provided by the U.S.-Earthquake Engineering Research Institute (EERI) as part of the 2010 Chile Earthquake reconnaissance efforts. The authors would like to acknowledge the Network for Earthquake Engineering Simulation (NEES) which provided the instrumentation and data acquisition equipment for the field study as part of the NEES-SHARED USE program. We would also like to acknowledge the help of a team consisting of local and international professionals and students without whom this process would not have been made possible at the speed and organizational management at which it was completed: Dr. Robert Nigbor (NEES@UCLA), Dr. Alberto Salamanca (NEES@UCLA), Matias Chacon, Javier Encina and Joao Marques (currently/formerly at Pontificia Universidad Cathólica de Chile), Dr. Marc Sereci (Digitexx) and Aditya Jain (ON Semiconductor), Aziz Akhtary (formerly at CSUF) and Pedro Quezada (formerly at University of Los Andes, Santiago). References [1] Los Angeles Tall Building Seismic Design Council – LATBSDC. An alternative procedure for seismic analysis and design of tall buildings located in the los angeles region.
[8] Bradford SC, Clinton JF, Favela J, Heaton TH. Results of millikan library forced vibration testing, technical report: Caltech EERL-2004-03. California Institute of Technology; 2004. [9] Doebling SW, Farrar CR, Prime MB, Shevitz DW. Damage identification and health monitoring of structural and mechanical systems from changes in their vibration characteristics: A literature review. Los Alamos National Laboratory report LA-13070-MS; 1996. [10] Celebi M, Sanli A, Sinclair M, Gallant S, Radulescu D. Real-time seismic monitoring needs of a building owner – and the solution: a cooperative effort. Earthquake Spectra 2004;20:333–46. [11] Naeim F. Performance of 20 extensively-instrumented buildings during the 1994 Northridge earthquake. Struct Des Tall Build 1998;7(3):174–94. [12] Goel RK, Chopra AK. Period formulas for moment resisting frame buildings. J Struct Eng, ASCE 1997;123(11):1454–61. [13] Goel RK, Chopra AK. Period formulas for concrete shear wall buildings. J Struct Eng, ASCE 1998;124(4):426–33. [14] Earthquake Engineering Research Institute (EERI). The Mw 8.8 ChileEarthquake of February 27, 2010. EERI Special Earthquake Report. June 2010.